Search Results

Is it necessary to account for the eccentricity on a WT section used as a compression or tension member? That is, can AISC Steel Construction Manual Table 4-7 be used for a WT compression member connected at the flange only?

Is the variation of the Uniform Force Method found in the 14th Edition AISC Steel Construction Manual section titled "Analysis of Existing Diagonal Bracing Connections" applicable to new construction?

Reinforcement may be required due to shear lag in heavily loaded HSS tension member connections. Does AISC have a published approach for the design of reinforcement for square HSS at slotted HSS-to-gusset connections?

Can an ASTM F1554 Grade 105 anchor rod be hot dip galvanized?

I would like to make some welded modifications to an existing steel structure that was constructed with ASTM A7 steel. Is welding to A7 steel acceptable?

How should I design the fillet weld for the load condition shown below?

AISC Specification Table J2.5, Available Strength of Welded Joints, lists different PJP groove weld strengths in compression for "connections of members designed to bear other than columns" and "connections not finished-to-bear." Please explain what is meant by that and why the "connections not finished-to-bear" have a higher strength.

I have not been able to find information on which manufacturers make an appropriate paint for either a Class A or a Class B faying surface for slip-critical bolted connections. Where can I find this information?

I am trying to determine the design moment for an FR beam-to-column moment connection in an ordinary moment frame. AISC 341-05 Section 11.2a specifies the required flexural strength as the lesser of 1.1RyMp or the maximum moment that can be developed by the system. How do I determine the maximum moment that can be developed by the system?

     I have a contract in which the connections have been specified as Option 3 connections per the 2010 AISC Code of Standard Practice. That is, the connections are to be designed by our engineer with sealed calculations provided as the substaining connection information. Shears and moments (formoment connections) were provided in the design drawings.

     We submitted representative samples of the substaining connection information, which showed the typed of connections (shear tabs and bolted flange-plate moment connections) and teh calculations we planned to submit to justify our connections. The final substaining connection information was consistent with the representative samples, and we submitted it to the Structural Engineer of Record (SER) with the shop nd erection drawings as a part of the approval process as given in Section 4.

     The returned shop drawings were marked up to require one additional row of bolts in each moment connection flange plate, even though the claculations we submitted demonstrate compliance with the requirements on the design drawings and in the AISC Specification, and recommendations in the AISC Manual.

     When we asked why the additional bolts were necessary, the SER responded that they wanted more strength in the connections. We objected because of the additional costs of revising uor shop drawings and changing the connections. the SER responded that the 2010 AISC Code of Standard Practice gives them final authority to require this, per Section 4, and additional compensation is not required. Is this option consistent with the intent of the Code?

I am designing a Four-Bolt Stiffened (4ES) Extended End-Plate Moment Connection in a Special Moment Frame. I have been designing according to AISC 358 Chapter 6. Table 6.1 has a limitation for the width of the end-plate of 10¾ in. maximum to 10¾ in. minimum. Am I only allowed to use a 10¾ in. wide end-plate?

The 13th Edition AISC Steel Construction Manual discusses the analysis of existing diagonal bracing connections using the Uniform Force Method. How is objective function given on pages 13-10 and 13-11 used in distributing the moment in the bracing connection and how is the Lagrange multiplier calculated?

I noticed that there is a tw3 term in the numerator and denominator of AISC Specification Equation J10-7 for the web sidesway buckling case where the compression flange is not restrained against rotation. Since these terms cancel out, why are they included in the equation? Is it true that the web sidesway buckling strength is independent of web thickness?

On page 3-28 of the AISC Seismic Design Manual, the gusset-to-beam force was calculated assuming only a single line of weld. Is this a conservative assumption? Figure 3-7 on page 3-34 indicates a double-sided weld.

AISC 360-05 Section I3-2a seems to imply that if h/tw is less than or equal to 3.76√(E/Fy), then the condition of the stresses under construction loading need not be combined with stresses induced on the post-composite section. The 1989 AISC Specification required a check of MDL/Sbeam+MLL/Seff ≤ Fa. Is it correct that when using AISC 360-05 Section I3-2a, the superposition of construction stresses in the beam and post-composite stresses on the transformed section need not be checked?

When reviewing the width-thickness ratios of elements in a custom shape, I found the term "NA" under λp in ANSI/AISC 360-05 Table B4.1. Does that signify that for this case, the shape having this element may be considered compact until the width-thickness ratio of that element reaches the limit defined by λr?

Section 4 of the 2009 RCSC Specification allows snug-tight shear bearing joints with tension as long as the tension load is "static." What is the definition of a "static" load? Are wind or seismic loads considered to be static loads?

Can I fill an HSS or steel pipe with concrete to serve as fire protection?

Our office practice for concentrically braced frames is to include a redundant moment frame within the braced bay. We typically ask for the following with respect to the lateral system connections:

• Braced connections to be designed for the full tensile capacity of the brace (0.6AgFy)
• That the beam-to-column connection to be designed for the full elastic capacity of the beam (0.66SxFy)
• That the scheduled beam shear be increased by (0.66SxFy) ×2/(length of girder)
  

Can you please comment on this practice?

I am analyzing steel roof trusses for a building constructed in 1928 with double angle webs and chords. I believe that the material was supplied from Bethlehem Steel. The top chord appears to be 2L5x3 with a thickness of either 9⁄16 or 5⁄8. Do you have any shape properties for this size angle? My manuals only go back as far as the 6th edition, and neither the AISC Iron and Steel Beams 1873 to 1952 nor AISC Design Guide 15 include properties of angle shapes.

I am also looking for information on structural steel and rivet properties for that timeframe.

Is it possible to add strength to an existing composite beam by adding mechanical anchors (expansion bolts) from below? The idea is to drill through the flange and install the anchor through the flange and into the concrete slab above.

I am working on the design of box sections in shear. According to Chapter G of the AISC Specification, the coefficient Kv should be taken equal to 5 for box sections. The commentary indicates that this is because the elements are restrained. Is Kv = 5 appropriate for each axis of a box shape?

Section J2.2b of the AISC Specification states that the maximum size of a fillet weld permitted along edges for material 1/4 in. or more in thickness, shall not be greater than the thickness of the material minus 1⁄16 in., unless the weld is specifically designated on the drawings to be built out to obtain full throat thickness. Does this mean that the weld size can be larger than the thickness of the thicker part?

Is it required to use slip-critical connections for high-seismic applications?

I am trying to calculate the flexural-torsional buckling strength of a non-standard T-shape per Specification Section E4. How are the torsional constant and warping constant calculated?

Table 3.1 of AWS D1.1 lists the required preheat for welding based on the category of the base metal. When welding steels of different categories together, which category requirement do we follow?

If I purchase steel plate as a raw material for a project and the customer is quoted a price per pound for the product, is he required to pay for the remaining skeleton of the steel plate if it is not useable in the project? Example: I purchase sheets that are 4 ft by 10 ft and burn two pieces that are 44 in. by 58 in. from each piece. I have a skeleton left over that is not useable. Does the customer pay for 40 sq. ft of material or only the weight of the two pieces?

What yield strengths are typically available for HSS2x2x1/4 and HSS3x3x1/4?

What is the normal torque for ASTM A325 and A490 bolts used in non-slip-critical connections?

If an oxy-acetylene torch is used to flame cut the edge of a new steel beam or plate, what impact does that have on the steel? Is the strength of the steel affected? Is the edge distance or bolt hole spacing affected by this process?

For base plate anchor rod design, why is ASTM F1554 the preferred specification? Is ASTM A449 suitable? What is the reference for torque values of base plate anchor rods?

All of the equations for prying action include the symbol p (the spacing between bolt rows). However, if we have only one bolt row in a connection, there is no spacing between rows. If p = 0, it seems to invalidate equations related to calculation of prying action. Are there alternative equations when p = 0?

How do I calculate the shear center of a rolled shape?

In the October 2006 issue of Modern Steel Construction, a range for the support line to bolt line (a-distance) is given as 2½ in. to 3½ in. for single-plate shear connections. Does this necessarily mean a = 2 would not work properly? If the connection were a through plate passing through a 6-in. column could a = 2 in. be used?

We are modifying an existing structure that has wide-flange vertical bracing members that were designed using the 7th edition AISC Manual. The connections used friction-type bolted joints with oversize holes in web plate connections to the gusset on both edges. The AISC Specification requirements applicable to the 7th Edition did not require friction bolts to be checked for bearing, as is required today.

In general, the connections work for the slip-critical portion of the calculation, but many fail the bearing strength check. In order to prevent the connections from slipping into bearing, I’ve considered adding welds to the joint. Do you have any suggestions as to how to evaluate these connections?

What constitutes a lateral brace for a beam? Does properly attached roof deck act as a continuous brace for the compression flange?

I am designing a monorail beam, which is an S-shape with a cap channel. I’m having trouble determining the limiting width-thickness ratios for strong-axis bending per Table B4.1 of the AISC 13th Edition Manual. For strong-axis bending I am checking three components:

• channel web between two fillet welds per Table B4.1 Case 12
• channel web between channel top flange and fillet weld as stiffened elements per Table B4.1 Case 12
• S-shape beam flange as unstiffened element per Table B4.1 Case 2

Am I doing this correctly?

I am reviewing an existing built-up column. The section is singly symmetric (symmetrical about the weak axis). The column is subject to combined axial force and flexure about the strong axis. Does the web element for uniform loading fall under Table B4.1, Case 14 of the 2005 AISC Specification? While checking the limit states of flexural-torsional and torsional buckling, I am using Equation E4-5 for singly symmetric members. Is this the correct equation when the axis of symmetry is the weak axis?

Table 2-2 of FEMA 350 allows bolted flange plate (BFP) moment connections as prequalified moment connections for OMF and SMF in high-seismic applications. ANSI/AISC 358 makes no mention of this type of connection. Is the use of BFP moment connections still permissible in high-seismic applications?

I have been using the ASD 9th Edition Manual. I am trying to learn how to use the 13th Edition. I am having a hard time finding the allowable stresses for different members, such as tension members, compression members, and members in flexure just to name a few. Is the bending stress for flexural members still 0.66F_y and 0.6F_y, depending on my unbraced length? Where are these located?

We are doing a project involving inspections of truss bridges, most of which were built in the early 1900s and are connected together with gusset plates attached with rivets. We have not been able to locate any literature relating the diameter of the head of the rivets to the shaft diameter. Is there any reference material that denotes the relationship of the diameter of the head to the actual shaft diameter?

Is there, or will there be, an update of the 1997 Hollow Structural Sections Connections Manual? I am particularly interested in finding information pertaining to single-plate shear connections to HSS.

Section 7.2 of the AISC Seismic Provisions indicates that bolted joints must have Class A faying surfaces. Section 3.2.2(c) of the RCSC Specification for Structural Joints using ASTM A325 or A490 Bolts indicates that galvanized faying surfaces are designated as Class C. Does this mean that we are unable to use steel members that are galvanized in the vicinity of the connections, for high-seismic applications?

A note at the bottom of the Floor Plate Bending Capacity table on p. 2-145 in the 9th edition ASD Manual indicates that the loads are based on an extreme fiber stress of 16 ksi and simple-span bending. The 16 ksi allowable stress seems to be very conservative, assuming that the plates would likely have a yield strength of not less than that for A36 steel. What is the 16 ksi allowable based on?

AISC Specification Section J10.7 requires full-depth stiffeners at the “unframed ends of beams and girders.” What does this mean? Would an example be a girder bearing on a column with no beam framing into it at the column?

Why is there a requirement to check punching shear on the wall of an HSS column with a single-plate shear connection, but no similar check when connecting to the web of a W-Shape?

Are C_b values permitted in the design of HSS beams? Are C_b values greater than 2.3 permitted, in any case, in ASD? Is there an instance where C_b = 4.7 for an unbraced square HSS cantilever with a concentrated load at the end is justifiable?

I am trying to better understand when to specify F1554 vs. A449 for anchor rods. Table 2-5 in the Manual does not indicate a preferred material specification for high-strength anchor rod.  Is there a reason for this?  Is there a preferred material for anchor rods?

I have several design spreadsheets that are written around the ASD 9th edition Manual, and I was wondering if any of the new IBC or ASCE codes reference a certain manual, like the new 13th edition, or if the older edition is still acceptable to use for design purposes.

I have been questioned about calculations for a stability bracing member per AISC 360-05 Appendix 6, Equations A-6-7 and A-6-8. I can calculate the required brace stiffness, but how do I calculate the actual brace stiffness provided?

I am trying to better understand when to specify F1554 vs. A449 for anchor rods. Table 2-5 in the Manual does not indicate a preferred material specification for high-strength anchor rods. Is there a reason for this? Is there a preferred material for anchor rods?

I’m wondering how to analyze a W-shape beam at an inflection point in terms of the C_b value. I understand that the inflection point cannot be considered a braced point, so when you calculate the C_b value some moment values will be negative and some moment values will be positive.  The Specification says that C_b is permitted to conservatively be taken as 1.0. Is this a requirement? If not, then which one is true in order to get a C_b value?

Use the absolute values for M_a, M_b, M_c and M_max across the entire unbraced section. (This would include positive and negative moments.)

Use the absolute values for M_a, M_b, M_c and M_max from one brace point to the inflection point. (This includes only negative or positive moments.)

Design examples 3.10 and 3.11 in the Seismic Design Manual, take the stress for the welds as the greater of f_peak or 1.25f_avg. Is this design practice specified somewhere in the Manual or some other publication?

Are C_b values permitted in the design of HSS beams? Are C_b values greater than 2.3 permitted, in any case, in ASD? Is there an instance where C_b = 4.7 for an unbraced square HSS cantilever with a concentrated load at the end is justifiable?

I have heard that rigid-frame beams in multi-story construction should not have welded studs applied to make them composite beams. Can you please give me a reference where this is stated?

Table 10-1 Page 10-23 of the Steel Construction Manual for All-Bolted Double-Angle Connections lists the available ASD capacity at 32.6 kips for 1/4-in.-thick A325N bolts. However, my calculation for angle shear rupture shows a value of 33.7 kips. If the hole diameter is changed from 13/16-in. to 7/8-in., the value is 32.6 kips, per your table. Is a 7/8-in. hole size assumed for a 3/4-in.-diameter bolt, and if so, where is this stated? Otherwise, how is the 32.6 kip value obtained?

AISC “Basic Design Values 1” (laminate card, copyright 2005) shows bending about weak axis = 0.9F_yS_y (ASD). Please verify the coefficient is 0.9 and not 0.75.

The last paragraph on 10-103 of the 13th edition AISC Manual states that this design procedure permits the column to be designed for an axial force without eccentricity. Does this also apply to girders used as the supporting member? In other words, if this design procedure is followed for a beam-to-girder connection, does the girder need to be designed for an eccentric load from the extended single-plate beam connection?

Could you help me understand the connotation of the "Compactness Criteria for Angles Table" shown on page 1-47 of the 13th edition Manual?

Is an IMF connection required to be stronger than an OMF connection?

Table B5.1 in the 9th edition Manual incorporates a kc factor for defining the Compact/Non-Compact behavior of flanges for I-shape welded beams in flexure. Why is this factor used and is there any update to this provision?

Is it okay to heat and bend a structural bolt that was placed in concrete in the wrong place? The contractor placed the bolts anywhere from 1.5” to 3.5” out of alignment, then heated the bolts and bent them like a snake and told me that they would be okay. I do not feel comfortable with this assurance. I did find one article that says to look in the 9th edition of the AISC Manual on page 4-4, but I do not understand what this means. Could you explain this in more detail?

My understanding is that we are able to consider all beams in a building as restrained for fire protection requirements, regardless of whether they are part of an interior bay or an end bay. Can you provide me with insight into the accuracy of this statement?

Section 6.4.4 of the AISC Code of Standard Practice includes the following sentence:

“For the purpose of inspection, camber shall be measured in the Fabricator’s shop in the unstressed condition.”

What does the term “unstressed” imply? Does the “unstressed condition” include or exclude the dead weight of the beam?

Section H1.1 of the AISC Specification states, “...Iyc is the moment of inertia about the y-axis referred to the compression flange.” Could you please expand on the definition of Iyc?

What does the rts listed in Table 1-1 of the AISC Steel Construction Manual represent? How is this function calculated?

When designing a channel for flexure, I am somewhat confused regarding the allowable/available moments that are published in the 13th edition Manual. 

I came to the conclusion that all of the channels now have an allowable bending stress of 0.75F_y as opposed to the older 0.66F_y.  Am I correct in assuming that this is what the new allowable stress is for channel beams?  Can I also get an explanation as to why the sudden increase in allowable stress has been made to be 0.75F_y?

When were A36 and A6 steels in general use for building construction?

Is a frame with a kick brace considered a special braced frame?

What type of structure requires a dynamic analysis?

Per the discussion on page 10-151 of the 13th edition AISC Steel Construction Manual, the maximum beam-web thickness is a function of the maximum root opening and the angle of skew in a skewed single-plate connection. Why?

I have also heard that some proprietary connections have been submitted for prequalification. Is there a location where I can find a list of the connections that have been approved since the publication of AISC 358-05?

My question pertains to the design of column web reinforcement for directly welded flange moment connections.

When local yielding of the column web occurs, is it acceptable to use web doubler plates in lieu of a pair of transverse stiffeners to provide for the additional material necessary to exceed the design strength requirements? The reason I ask is because all of the AISC design examples, and also the software that we use for connection design, always provide for transverse stiffeners instead of doubler plates when local yielding of the column web is an issue. 

On one of my projects, the fabricator is asking to use thermally cut bolt holes. He is citing Section M2.5 of AISC 13th edition Steel Construction Manual. That section states that thermally cut holes shall be permitted with a surface roughness not exceeding 1,000 min.  

What are the advantages and disadvantages of thermally cut holes?

How does roughness of surface come into the picture for cutting the holes?

A steel subcontractor on our project says he has a U.S.-made A500 Grade B round HSS. I believe that this material only comes in square and rectangular shapes. Can you confirm the availability of these shapes?

How are ASTM A325SC bolts designed and installed? In past projects we typically used A325N and we are now using A325SC. Do both of these types of bolts need to be checked for bearing?

We have a 120-ft-long, 12-ft-high truss made up of wide-flange beams for the upper and lower chords. The truss will be fabricated with a 5-in. camber. At the worse loading condition, the truss will still be 1 in. above flat. How is the connection made from the member end to the column? Will the fabricator detail the member so the bolt holes in the cambered member are vertical and in line to the connection plate or angles holes, or do they offset the bolt holes? If the truss is to rotate downward 4 in., should oversized holes be used in both the plate and member to allow for rotation?

Can anchor rods be welded to the reinforcing bars to be embedded in the slab to be cast? I cannot find anything in the codes in reference to this subject. Can you help?

Where can I find information pertaining to compatibility when using different types of base metal and fasteners in a connection?

Where can I find information for the allowable amount of beam sweep in an erected building?

Section J10 of the AISC Specification addresses double-concentrated forces. Could you please explain the meaning, or give an example of a double-concentrated force as it applies to this section? 

In discussions with several engineers, I am hearing it said that the 13th edition is forcing engineers to abandon the ASD method, and to conform to the LRFD method.

Two things that I remember from this year’s AISC seminar on the 13th edition are:

• The LRFD strength is equal to 1.5 times the ASD strength, and
• either approach can be used, and the designer just has to remain consistent with the chosen method during the calculation.

I need some help understanding the Block Shear Equation (J4-5) in the 2005 Specification. For block shear strength:

R_n = 0.6F_u A_nv + U_bs F_u A_nt ≤ 0.6F_y A_gv + U_bs F_u A_nt

The left side of the equation must be less than or equal to the right side. For ASTM A36 steel, F_u = 58 ksi and F_y = 36 ksi, which means that 0.6F_u will always be greater than 0.6F_y. The net area in shear, A_nv, is smaller than the gross area in shear, A_gv, but not enough to overcome the difference between 0.6F_u and 0.6F_y.

As an example, consider a 3/8-in. plate with 1.5-in. edge distance and 6 rows of 3/4-in. bolts in STD holes at 3-in. spacing.

A_gv   = ( 3/8-in.)(5 x 3-in. + 1.5-in.)

= 6.19 in.²

A_nv   = A_gv – hole area deduction

= 6.19 in.² – ( 3/8-in.)[5.5 holes x (13/16-in. + 1/16-in.)]

= 4.39 in.²

From this, 0.6F_u A_nv = 153 kips, and 0.6F_y A_gv = 134. That is, this shows 0.6F_u A_nv > 0.6F_y A_gv.

So where did I go wrong? When will 0.6F_u A_nv ever be less than 0.6F_y A_gv?

Section 2.9.1 of Design Guide 1 contains the following statement:“When designing anchor rods using setting nuts and washers, it is important to remember these rods are also loaded in compression and their strength should be checked for push-out at the bottom of the footing.”

How does one go about calculating the push-out of the anchor through the bottom of the footing?

What is the AISC position on use of LRFD or ASD design? It appears that the equations in the 2005 AISC Specification can be utilized in the LRFD or ASD method by multiplying by the Ω factor or dividing by the φ factor. Is this correct, because people are telling me one cannot use the ASD method anymore? What about the IBC—do you know if they specifically require the use of LRFD?

In the Specification for Structural Joints Using ASTM A325 or A490 Bolts, in the last paragraph of the Commentary to Section 4.1, it states that “snug-tightened installation is not permitted for these fasteners in applications involving non-static loading” for bolt assemblies in direct tension.

We have drag struts with relatively small seismic loads, compared to the gravity loads. In such cases, there will be no “net” non-static loading whatsoever. As such, is a snug-tightened joint acceptable?

I have questions pertaining to Equations (C-F4-3) and (C-F4-4) shown in the Commentary to the 2005 AISC specification. What is the significance of α − β_x? What is the F_yr being used in the calculation of L_r? It is not defined.

One of the recommended methods for installing bolts is the turn-of-nut method. The RCSC specification indicates that turn-of-bolt can be used if it is impractical to use the turn-of-nut method. How is the turn-of-bolt method different than the turn-of-nut method?

Please confirm that h/t_w ratio in footnote [a] of Table B4.1 (page 16.1-18 in the 13th edition manual) is the height of the web (i.e., clear distance between flanges) over the web thickness (see Case 2). We just want to verify that this ratio should be used, and not the b/t ratio of the flange.

Does AISC provide any written material that compares the properties of old designated beams sizes, such as 16B26, to the current properties of a standard wide-flange (W shape)?

I have a question on AISC 341 Section 14.3 regarding the unbalanced earthquake load acting on beam due to a buckled brace. We have a two-story OCBF where the V and inverted V braces meet at the second floor beams, forming a two-story X. The braces above and below the beam tend to balance each other (opposing forces). We are not clear if this requirement applies to this configuration. Does it apply only for Chevron-type configurations where the braces are located only below (inverted V configuration) or only above (V configuration)?

I am looking into the difference between “restrained” and “unrestrained” ratings for a steel-framed building. I have read the AISC engineering FAQs at www.aisc.org/faq, as well as the ASTM E119 and other literature, and all seem to point me in the direction that a standard steel building is classified as restrained.

    That said, we have had some discussion in our office that the restrained classification depends on the continuity of the structure. While I understand that continuous beams spanning over more than two supports will offer more rotational restraint that a simple shear connection, are the simple shear connections enough? Could a single-bay structure (with simple shear connections) be considered restrained? Are there any special restraint requirements for perimeter beams? If one part of the structure becomes unrestrained, does that mean that the entire structure must be classified as unrestrained?

In preparation for an ICC bolting examination, one of our technicians found a question asking what structural group number a W14×426 member was. We found the answer in the AISC 9th edition ASC manual, but not in the 13th edition Steel Construction Manual. The ICC test now references the newer manual, and I was wondering if this table is included in it, or if it was left out intentionally?

We have encountered a situation where it appears that the carbon steel framing is delaminating, for lack of a better word. I have searched the Internet, but have not found many articles or data concerning this specific issue. Do you know of the correct terminology to use when the steel is separating in layers?

Do horizontal bracing members need to meet the requirements of Sections 13.2a and 13.2d of the 2005 AISC seismic provisions? It is not clear whether “bracing” applies to horizontal and vertical, or just vertical bracing.

What is the best way to specify simple shear connection reactions on the design drawings? On a non-composite steel framing system, we typically specify that “All beam connections shall develop the full uniform load capacity the member can carry...” The connection designer can then easily obtain a design load using the AISC allowable load tables. This method also ensures that the connection will not be the limiting design element. We have found this procedure to be an efficient method to specify design loads for typical framing.

    We would like to specify composite beam connection design loads in a similar manner. Is there a design aid available that would allow a connection designer to easily obtain the maximum uniform load for a composite beam?

    Do you have any other recommendations to efficiently provide simple shear connection design loads for composite framing?

It is with great sadness we report that David T. Ricker died on February 22, 2008 in Payson, Ariz. Dave was a longtime member of the AISC Committee on Manuals and Textbooks and a frequent contributor of articles, papers, and answers to questions about steel design and construction, including in this feature, Steel Interchange. He authored several professional articles that are still provided with great frequency in response to questions received by AISC. To honor the memory of this friend of the steel industry, Dave is the focus of this month’s Steel Interchange.

It is a general rule that welding on an existing structural member is not permitted unless provisions are made to unload the member first (for example, if the member is being reinforced) and that the weld must not degrade the properties of the material. Is there a written reference that discusses this from both a code perspective and a practical approach?

A typical lifting beam or strongback in the materials handling, crane, and rigging industry takes the form of either a horizontal or wide-flange beam, with padeyes top and bottom at both ends. The lifting wire rope bridle with two legs at about a 45° angle attaches to the top padeyes, and the supported weight attaches to the bottom padeyes.

    The wire rope bridle induces both compression and bending moment in the lifting beam. Again, there is no lateral support. What analysis would be used to solve for the safe lifting capacity of this form of lifting beam?

How is a welded double-angle connection designed when the [wider] double angles are connected to the [narrower] flange of the column and welded on the back side of the double angles? This may be necessary when the column flange is short.

How does the AISC Code of Standard Practice address the possible tolerance for vertical and horizontal alignment of a crane rail in a mill-type building?

During bridge repair, rivets are often removed and replaced with ASTM A325 or A490 bolts. Is there a standard procedure written for the removal of rivets and resizing of the fastener hole? If the base metal is going to be reused, I would think that it would be very important not to damage or overheat the base metal around the fastener hole. This base metal could be a multiple build-up of two-, three- or four-plys. Should these rivets be removed with a machine or cutting torch? Rivets are pressed in when newly installed; should they be pressed out? What preparation should be taken to remove and rework a riveted connection?

The elastic moment strength of a beam listed in Table 3-6 of the 13th edition AISC manual seems low as compared to the 9th edition ASD manual and the 3rd edition LRFD manual. As an example, for a W16×77 with Fy = 50 ksi:

    13th edition,Table 3-6, pp. 3-60

       ASD: M_r/Ω = 234 kip-ft

       LRFD: ΦM_r = 352 kip-ft

    9th edition ASD, pp. 2-11

       ASD: M_R = 369 kip-ft

    3rd edition LRFD, Table 5-4, pp. 5-62

       LRFD: ΦM_r = 405 kip-ft

Could you tell me why there are such differences?

1. When the cuts for an RBS are determined by a, b, and c (identified in AISC 358-05, Prequalified Connections for Special and Intermediate Steel Moment Frames for Seismic Applications), does this cut section need to develop the maximum moment as required for that member?
2. Bracing as per AISC 358-05, Section 5.3.1 ,(7) is required for RBS beams. If there is a steel decking floor system with shear studs on top of the RBS beam, can this flooring system be considered an adequate form of bracing for this beam?

When considering nut installation on a threaded anchor rod, how is the pretension specified on the drawings? The desire is to prohibit loosening of the nut under load reversal. I see the recommendations for anchor-rod nut installation on pages 14-10 and 14-11 of the 13th edition AISC manual; however, the minimum bolt pretension force shown in Table J3.1 does not seem to apply. Is the required tensile force still 0.7F_u of the anchor rod?

The title of Section H1.3 of the 2005 AISC specification is “Doubly Symmetric Members in Single Axis Flexure and Compression.” This title indicates that this section can be used for doubly symmetric I-sections and rectangular (or round) HSS as well. Bending can be either in the major axis or the minor axis. After reading the Commentary, my understanding is that this section is intended for doubly symmetric I-sections subjected to major axis bending only. Is section H1.3 applicable to HSS?

I am searching for information pertaining to r_t. I am looking to see if there are any equations that can be used to calculate this value. I have looked through a couple of text books but have not found an equation for this. Can you point me in the right direction?

What is the difference between LRFD and ASD design?

The Brockenbrough and Merritt text referenced in Part 2 of the 13th edition AISC manual indicates that “For special elevated-temperature applications in which structural steels do not provide adequate properties, special alloy and stainless steels with excellent high-temperature properties are available.” Can you direct me to publications on the properties of these special alloy and stainless steels?

In considering a brace-to-gusset connection for a SCBF, it is my understanding that the Uniform Force Method is generally the most economical method for determining gusset size and welds (13th edition page 13-3). However, I noticed that most of the examples in the new Seismic Design Manual use the Whitmore section (pages 3-58, 3-66, etc.). Are there any advantages and/or restrictions in using the Whitmore section versus the Uniform Force Method? Is there any reason the Whitmore section was used in the Seismic Design Manual versus the Uniform Force Method?

I recently attended the AISC seminar Design Steel Your Way (www.aisc.org/seminars) and received the AISC manual and companion CD. In a project I am working on we have a shape that we cannot locate in the historical shapes database that was included on the CD, nor in any texts that we currently have. I was wondering if you might be able to shed some light on some of its dimensions and properties.

     The building was built by the government in 1951. On the drawings the shape is called out as a 12WF19. We are looking to do some analysis on this structure and wondering if you could direct us to where we can find the properties?

AISC 341-05 requires 1.1R_y for the design of some connections and R_y in other places. Why is this so? Should it not be only 1.1R_y, since I understand the 1.1 is to account for the increase in strength due to strain hardening under cyclic load? If this is the case, all elements would be subject to cyclic loading in a seismic event, and hence 1.1R_y should be applicable everywhere.

We are examining an existing building constructed roughly in 1919. The columns in the building consist of double-angle flanges with lattice steel webs. We are trying to determine allowable loading for the column but have been unable to find any documented information. Any help would be appreciated.

I am working on an addition to a 1922 steel-framed building. When I check the columns using ASD (unreinforced concrete encased, F_a = 18000 – 70L/R), many are overstressed. However, the results appear to be better using LRFD.

Is there anything in the AISC specification that would require that we use ASD and not LRFD to check the existing structure? Is it reasonable to apply today’s load and resistance factors to historic steel?

Is there a minimum height requirement for studs used in the design of composite beams?

Where are “crane runway horizontal forces” addressed in the current AISC specification? This was formerly covered in Section A4.3 of the 1989 ASD specification. If it is not addressed in the new specification, where do I need to go to find these requirements?

What is the maximum length permitted between expansion joints in a steel building?

In the 3rd Edition AISC LRFD manual, footnote "e" in Table J3.2 requires that a 20% reduction is taken on bolt patterns in tension splices where the bolt pattern length measured parallel to the line of force is greater than 50 in.

Is any tension connection considered a splice? I am evaluating diagonal tension members in a truss that connect to the top and bottom chords as well as vertical members. Does this reduction apply?

I am doing a seismic retrofit for a structure designed and built in 1931. I guess it must be A7 steel. Could you let me know what values F_y, F_u, and R_y should be used?

Can you help me clarify the design of a single unequal leg angle in bending utilizing the new AISC specification? A very common situation for longer span masonry brick openings is to use an L6x31/2x5∕16 LLV. Do you have an example or some clarification on the design?

I understand that the torque required for bolt installation in a slip-critical joint depends on the bolt class, the bolt diameter, and the faying surfaces properties. Can you tell me what equation to use to determine the torque required developing the required tension in the bolt?

I am relocating a structure where there is not proper documentation of existing conditions, and I am unable to determine the required length of the structural bolts.

We have the connection drawings of the columns and beams, and the thickness of the splice plate. I want to know what should be the length of the bolt and length of the threaded portion, taking into consideration the height of the nut and the thickness of the washer.

I have an H-shape in bending about the weak axis. Chapter F, Section F2 of the AISC manual notes, "Lateral bracing is not required for members loaded through the shear center about their weak axis, or for members of equal strength about both axes." Do we still need to consider the lateral torsional buckling effect if the beam is bent about the weaker axis? Based on my understanding, lateral bracing prevents torsion and lateral deflection only. What is the exact relation between "lateral bracing to prevent torsion" and "lateral bracing to prevent lateral torsional buckling"?

A fabricator has recommended providing member end forces (for beams) and axial member forces (for braces) for economical connection design. What are "transfer forces" with regard to connection design, and how does one calculate these? What is the rationalization as to why member end forces acting transverse to the longitudinal axis (in the weak axis direction) are ignored in connection design?

I understand that there is a provision that will allow me to not design steel connections per the AISC Seismic Provisions if I use an R value of 3. Can you tell me where I can find this information?

Could you suggest a resource where I can obtain mill pricing on structural shapes and information on availability? Could you also suggest a resource where I can obtain list of domestic steel distributors and major fabricators?

What is the minimum allowable bend radius for 1.5-in.-  and 1.75-in.-diameter bars in ASTM A572 Grade 50?  When hot bending is required, are there code provisions for the temperature required?

I was informed by my engineering team that a W24x160 beam is commercially available. Can you tell me by whom such a beam is being produced?

I remember seeing a download that shows the historical steel specifications used in industry—for example, what years ASTM A7 steel was used. Can you help?

Table 3-2 in the 13th edition manual has a column showing value of BF; however, the symbol list does not include the term BF. General Nomenclature (Index) indicates that BF is factor that can be used to calculate the flexural strength for unbraced length L_b between L_p and L_r. How is that done? Table 3-2 also lists M_rx. How was this calculated?

Does the limit state of yielding apply to an unbraced tee flexural member with stem in tension? The compression flange is non-compact.

I am concerned with a fabrication shop that is using E70C-6M. When inspecting heavy columns, I noticed the welds at the gusset plate have little or no penetration into the base metal. My wire book is saying the wire is good on carbon steel up to 70 ksi. Is the gas-shielded metal-cored wire, which I see as having little penetration, acceptable on structural steel because the drawing calls for the use of E70XX filler metal?

I am trying to calculate the I_x values found in Section 1 of the 13th edition AISC manual. The numbers that I get are close, but don’t quite match what is printed. Are you including the fillets in the calculation? What shape do you assume they are? Where is the thickness of the flange equal to t_f when shapes have an inner flange surface slope? Is it at the location half way between the face of the web and the end of the flange?

When AISC gave a seminar on the new code, they gave out laminated “short-cut” cards that identify the ASD tension capacity of a PJP groove weld as 0.32F_EXXA_w. I am having trouble coming up with the origin of the 0.32 factor. For ASD I have always used an allowable shear stress of 0.3 times the nominal tensile strength of the weld metal. Can you explain the difference?

Upon reviewing a condition I have encountered of an unequal-leg single angle bent about a geometric axis (X-X in this case) with no restraint against lateral-torsional buckling along its length, I do not see in AISC specification Section F10-2 any subsection that encompasses this condition. Is it advisable to assume that since this condition is not explicitly covered with a definition for M_e, that there is no case in which LTB will cause the failure of an unequal-leg single angle bent about a geometric axis? If not, which equation would be applicable to determine the LTB criteria for this condition? 

AISC 341 lists only Class A surface preparation for bolted connections that are part of the SLRS. This would appear to eliminate galvanizing as a corrosion control method, since, when roughened, it is considered a Class C surface. Is this correct?

What and where are the requirements for edge distance for the recommended maximum-size holes for anchor rods listed in table 14-2?

We are renovating a job dated 1914. Beams are called out as 9x21. Is this an "I-shape," 9 in. deep at 21 lb. per ft?  Is there a source for the old beam properties?

Is it acceptable to design single-plate connections for shear and axial loads? If so, what are the design criteria for the axial load?

When installing the light gauge framing on a special moment frame with RBS connections, four shot pins were inadvertently installed into a beam flange in a protected zone. What criteria can we apply to determine if this exceeds an acceptable level? And if it does, what repairs are available to us? 

What is the reason for the smaller gross areas of standard pipes in the new (13th edition) steel manual?

We are considering the use of 65 ksi steel for building columns. Can you tell me what the availability would be for 65 ksi W-shapes? We are currently looking at sizes ranging from W14x61 to W14x605—the whole range!
    Additionally, the project (located in Chicago) is to be LEED certified, so we would be looking for steel manufacturers within a 500-mile radius. Do you have any recommendations for steel suppliers that would be able to fulfill this requirement and supply 65 ksi W-shapes?

Where can I find the current design criteria for plate girders? What used to be Chapter G in the 1989 ASD specification doesn’t appear to be in the new 2005 AISC specification.

Is it required to have evidence of traceability of heat numbers for clip angles, bar sections, or plate from inventory?

Why is the maximum size of a fillet weld limited by the thickness of the thinner plate?

I have a multi-story building constructed circa 1970 that is being considered for re-use as a senior center. Should the structural engineer be required to check the bolts for conformance with modern criteria?

Does one use the 1.25 multiplier over the bolt shear values for the extended configuration single-shear plates? Per p. 10-103 in the 13th edition AISC manual, the 1.25 multiplier is used to determine the moment strength of the bolt group. But when you are calculating the bolt shear strength, is it correct to use the bolt shear values directly from the tables and multiply these by the "C" factor for the eccentricity?

Is there a minimum number of threads that a nut needs to be engaged on a bolt?

For a steel-framed project with concrete slabs on metal deck, I know that the AISC Code of Standard Practice sets the tolerances for the steel, but what typically defines the tolerances for the top of concrete slab on metal deck? Normally, the concrete subcontractor uses the ACI "F" number criteria from the cast-in-place section of the project specifications, but is this correct if one just references the basic ACI and AISC standards without additional project specifications?

Section 14.2c of the 1997 version of Seismic Provisions for Structural Steel Buildings does not allow tension-only bracing. The 2005 version states that tension-only bracing can be used, but not in K, V, or inverted V configurations.

     Am I reading this correctly? When did it change that tension-only could be used? I do see that there are many restrictions to this.

Please let me know whether a bearing-type connection can be pretensioned.

     If yes, should the pretension (70% of the proof load of the bolt) be taken into account and checked for combined shear and tension when using N- or X-type bolts?
     What is the purpose behind pretensioning bearing type joints?  Is the pretension to be in accordance with RCSC 2004?  Will prying occur in a pretensioned joint?

I am interested in purchasing a copy of the book Dimensions and Properties, Rolled Shapes—Steel Wrought Iron Beams and Columns, as rolled in the USA, Period 1873 to 1952. It was compiled and edited by Herbert W. Ferris, and it looks like the 9th printing was in 1983.

We need more information on Section J10.8 in the 2005 AISC specification. The basic question is: Where do the 25t_w and 12t_w parameters come from? What about when you have two stiffeners within 25t_w/2 of each other? Are the strengths additive? Is there any provision by which the 25t_w limit can be increased?

I am designing a building with vertical siding, channel girts, and sag rods. I was told that the industry standard is to consider the channel braced at the sag rods. How is this possible, since the sag rods are neither at the compression flange nor able to act in compression? Is this the appropriate way to look at bracing the channel, or is there a more appropriate method?

Do you have a good response to engineers who forbid the use of the AISC 13th edition for connection design on their projects? I am running across a few engineers who take the position that since the IBC predates the 13th edition, the 13th edition cannot be used.

How does AISC Certification for fabrication of steel building structures relate to the IBC 2003 Paragraph 1704.2.2, Fabricator approval?  If a fabricator is AISC Certified, can I assume they qualify for the special inspections exemption under IBC?

Should anchor rods be designed for base shear forces only, not amplified loads, using the Ω factor and not the vertical or horizontal component of the R_yF_yA_g force of the brace?

We do a lot of one-story commercial buildings where the base shear forces are quite small, and we are being questioned by fabricators, contractors, architects, etc. over huge gusset plate connections that result from the R_yF_yA_g of the tube braces. The footing, anchor rods, and even column sizes seem disproportionate to the brace connection. Where does the force go if you ignore it in the design of these members?

We are under the 1999 Standard Building Code, which does not differentiate between special and ordinary concentric braced frames (except through referring to ASCE 7). This code refers to the 1997 AISC Seismic Provisions, which would technically allow us to ignore the seismic provisions for one- and two-story structures. However, we have been trying to use the 2002 provisions to be more up-to-date. The 1997 provisions allowed for the brace connection for OCBF to be designed for load combinations 4-1 and 4-2, but we see that this was taken out in the 2002 provisions for OCBF, along with the one- and two-story exception. The SCBF still allows the connection to be designed for the maximum force that can be transferred by the system. If you design the foundations for the base shear and let that limit the system, then aren’t you basically cancelling out the requirement for the connection to be designed for the R_yF_yA_g force, which effectively puts you back at square one?

We have a steel building frame where we are using ordinary moment frames (OMF) with a seismic force resisting system defined as "Structural steel systems not specifically detailed for seismic resistance," R = 3. Since this category does not require compliance with the AISC Seismic Provisions for Structural Steel Buildings, which requires removal of backing bars for OMFs, I would surmise that we do not need to remove the backing bars. Is this correct?

On p. 7-19 of the 13th edition Steel Construction Manual, an equation is given for the pure moment capacity of a bolt group when the instantaneous center is at the center of the bolt group. Where does the 1.25 in the equation come from? What is the design capacity using LRFD?

For design of composite steel beams, is there a good reason to consider the use of a non-uniform spacing over the span? I have heard of placing more at the ends where the shear is higher. Would that be reasonable?

I am designing a new stair that will be supported on an existing structural steel composite floor system. The beams are designed very close to their limits, so I must reinforce them. When I attempt to add steel (WT, C, L) to the bottom flange of the beam, the percentage composite drops below 25%, and I lose the ability to consider the beam composite, which means I would have to add a lot more steel. Has AISC done any research on adding studs to a composite beam to increase its composite action capability in order to accommodate additional loads?

I am trying to design an unstiffened seated connection using Table 10-6 in the 13th edition manual. What is the definition of the required bearing length N_req (in.) shown in the table? Is this referring to the actual beam bearing length on the angle, and if so, why do the allowable load values decrease as the bearing length increases?

I recently attended a seminar on the new manual, which included a study book outlining the seminar and including examples. I have been trying to familiarize myself with the axial compression section of this book (Chapter 2). An example uses a W14x120 column checking P_a using equation E3-2. This result is then checked against the capacity listed in Table 4-1. The results come out the same. I did a second problem using a W12x72 following the same approach and am coming up with significantly different values. I am not sure that I am using r_x/r_y correctly. Can you explain the use of the r_x/r_y  factor in assessing the column capacity?

What is the required installation torque for each diameter of bolt?

If addtional external load is applied on a pretensioned bolt, is the load additive?

In Chapter G of the 2005 AISC specification, the h/t_w limit is 2.24√(E/F_y) for webs of rolled I-shaped members. This is a slight change from the 2.45√(E/F_y) from the 1999 LRFD specification. Why?

Ambient air temperatures in the northern regions can reach below zero routinely--say -20 °F. Is the AISC specification still valid?

Recently, I have seen many references to the "LRFD method" in connection with structural engineering. Can you please advise me on how I might become familiar with this procedure?

I am designing a 12-story concentrically braced steel frame building. The Seismic Design Category is C. I'm taking R = 3, so the AISC Seismic Provisions are not applicable. What's more economical: HSS field-welded braces or W-shape field-bolted braces?

Does the AISC specification distinguish between the uses of recycled steel versus that produced from raw ore?

I remember that AISC presented a series of seismic braced frame seminars last year. Are there any scheduled in the future?

When designing for wind moments with oversized holes in a flange plated moment connection, can the slip-critical bolts be designed for serviceability or must they be designed for strength based on the 2005 AISC specification?

We are designing a material handling system in a high seismic and high wind region, Seismic Design Category D. The system classification is Special Concentrically Braced Frames due to the height of the structures. Referring to the 2005 Seismic Provisions, is it acceptable to design the bent bracing as tension-only X-bracing?

There is no listing for yield strength of ASTM A307 bolts in Table 1-C of the 9th edition ASD manual. Why is this not listed, and what is the value for it? Also, what is the definition of proof load listed in the table?

I just saw the word "backstep" in reference to a welding note on a structural drawing. What exactly does this mean?

Where can I find the filler metal requirements for high-seismic design?

A two-story column on our project was erected approximately 1&7/8 in. out-of-plumb. The erector would like to heat one face of the HSS column and then cool it quickly in order to bend the column back to plumb. All other issues aside, what will this heating and cooling process do to the material properties of the column? It seems as though there will be some residual stresses in the column as well that may present a problem.

On one of my projects, it was reported by the inspector that holes in some of the column base plates were enlarged in the field to accommodate anchor rods that were misplaced. Some of the holes were enlarged significantly, and some of the plate edges were notched out around bolts. The columns are part of a moment frame, so the bases were designed for the lateral forces. Is it possible to repair the plates by welding an angle or plate on top with drilled holes to receive the anchor rods? If a new plate is added on the top, the anchor rods may not have adequate projection. The rods are A307 material. Is this weldable, or would a coupler be required?

In attempting to calculate the lateral-torsion buckling of double-angles per the AISC manual, I cannot find properties such as Iy and J. Are these tabulated in the 13th edition manual?

What is the definition of knifed connection? I have looked in several textbooks without any success.

The 2005 AISC Seismic Provisions for SCBF states:

Where the effective net area of bracing members is less than the gross area, the required tensile strength of the brace based upon the limit state of fracture in the net section shall be greater than.... R_yF_yA_g (LRFD).

For an HSS brace, F_y = 46 ksi; F_u = 58 ksi; R_y = 1.4; R_t = 1.3

The required tensile strength of the brace based upon the limit state of fracture is 0.75RtFuAe = 56.55Ae and the required tensile strength is RyFyAg = 64.4Ag. Setting the required tensile strength of the brace greater than RyFyAg results in Ae = 1.14Ag, which is not possible. It seems that you cannot stiffen the member to increase Ae without also increasing Ag. How can the requirement of 13.2b(a) can be met?

Has Table 3, "Allowable Load for Slip-critical Connections," which appeared in the 9th edition AISC manual, not been included in the 13th edition?

I had the impression that we could not use slip-critical bolts with extended shear plates because the flexibility of shear plate connection is achieved by the plowing of the bolts against the main material. Based on the 13th edition AISC manual, it seems we can use slip-critical bolts with the extended configuration. What makes the connection flexible in this case?

I cannot find the requirements for how to determine the thickness of a base plate subjected to a weak-axis column moment in AISC's Design Guide 1: Base Plate and Anchor Rod Design. Could you provide some references?

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I noticed that the bolt values have changed for A325 N bearing bolts from the 9th edition manual (7/8" A325 N bolt = 12.6 kips) to the 13th edition (Table 7.1 gives 14.4 kips). Is that true?

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On the contract drawings, the engineer calls for us to develop the column splice for 100% of the gross moment capacity of the upper column. Usually we use Table 14-3 from the 13th edition manual, and in this situation, Case VI, because the engineer requests bolted/welded. I am assuming this table does not incorporate the 100% requirements. If this is the case, can you lead us in the right direction to design for such requirements?

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What are the key advantages of choosing composite steel beam and steel deck with concrete slab instead of a non-composite system? Is it possible to specify that the shear studs be shop-welded to the beams prior to arrival on-site? How can overall quality of shear stud installation be ensured?

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What are AISC's recommendations regarding the use of single-plate beam-to-girder shear connections? Typically, we do not use this type of connection, but I was wondering if there is an article discussing the pros/cons of this connection type.

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What is the torsional unbraced length? Isn't this equal to the lateral unbraced length? If a simply supported W-shape beam is laterally supported at the top flange every 5 ft and at the bottom flange every 10 ft, is the lateral unbraced length 5 ft for the top flange and 10 ft for the bottom flange? Isn't the torsional unbraced length the same?

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I am designing a large retail store with moment frames. In one direction the wind governs and in the other direction seismic governs. For the flexible moment connection (Type 2 with wind), page 4-100 of the 9th edition manual mentions that the moment connection is to be designed for wind moment only, and it is assumed to rotate enough to be considered simply supported for gravity loads. Does this assumption also apply to moment connections for seismic load?

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What are the major differences between the ASD and the LRFD design methods in the current AISC specification?

What is the difference between ASTM A36 anchor rods and ASTM F1554 Grade 36 anchor rods? They both have the same tensile and yield properties.

My questions relate to panel zone shear and tension field action. The 2005 AISC Specification, Section G3 states that tension field action cannot be used in the "end panel." There is no definition given for end panel.

Is the end panel not intended to be a support zone from the bearing stiffener, if run up from the center of the support to a cantilevered or fixed end, where the deformations are constrained? Who originally contrived the end panel notion, and what reference could we use to determine the intent?

I'm designing a flare bevel groove weld between a rectangular HSS16x8x1/2 and an A36 plate. The load is parallel to the weld axis (tension and/or compression). Table J2.5 states "Tension or compression in parts joined parallel to a weld need not be considered in design of welds joining the parts." So, how do I design the weld?

I am working on design modifications to an existing power plant built between 1967 and 1970. I can reference back the steel member designations to the AISC Steel Construction Manual, fifth edition. Based on field observations, the building was constructed using standard beam connections. The general connection notes required the fabricator to design the bearing or friction connections (appears to be fabricators choice) for the forces indicated on the drawings or for the full member strength if no forces were given. The fabricator was to use rivets or high-strength bolts.

I have an old copy of the 5th edition AISC manual. I noticed connection strength tables for standard beam connections, but they are for rivets. Based on my field observations, the connections use 3/4-in.-diameter bolts. What were the shear and tension properties of high-strength bolts at that time?

In a recent project, a reference to "Technical Bulletin No. 3" was made when specifying steel material. Can you please advise where this bulletin can be found?

In the 9th edition AISC ASD manual, the value of r_t was tabulated, and one can refine it by using the St. Venant equation as given in Commentary. The 13th edition AISC manual tabulates values of r_ts. Are both values same? If not does one need to calculate the rt?

Section F9 of the AISC Specification covers tees loaded in the plane of symmetry. Equations F9-4 and F9-5 cover the stem being in tension or compression. I'm designing a member that is loaded in its axis of symmetry (the y-axis for a tee or double angle), which produces tension or compression in the stem. However, the formulas use I_y, the moment of inertia of the weak axis. Using the weak-axis moment of inertia for bending about the strong axis is counterintuitive to me. Why is the weak-axis moment of inertia used in these equations?

I am examining a set of documents from 1964. Which manual of steel construction might have been used for steel design in 1964? I see shapes listed as 14B22, 14B17.2, etc. Furthermore, the drawings simply state that "fs = 22 ksi." Was steel of this vintage ASTM A36 with F_y = 36 ksi or something else?

I have a question on an example in the 2005 AISC specification seminar. You apply Eq. E7-17 to obtain the equivalent width of the web. AISC specification Section E7.2(a) indicates that this equation is valid only for b/t ≥ 1.49(E/f)^0.5. The value of this expression is 1.49(29,000/28.2)^0.5 = 47.8. The value of b/t = 39.6, which is less than 47.8 and it would seem that Eq. E7-17 is, therefore, not applicable. No alternative to Eq. E7-17 is provided in the specification. What is the intention of the AISC specification in this situation?

Is a backing bar required when welding the beam bottom flange to the column flange on a SMF if the beam flange is tight to the column?

I noticed that the tension yield component of the block shear check is no longer included in the 2005 specification, Eq. J4-5. The new procedure is obviously simpler without the need for dealing with the tension yield component. Could you explain the change?

I am trying to understand the requirements for bracing connections in AISC 341-02 (the 2002 AISC seismic provisions). In Sections 13.3a and 14.2, the required strength of the connection is the expected tensile strength of the brace, R_yF_yA_g. How can I meet that strength requirement for tension in the connected elements? According to Sec. 13.3b:

The design tensile strength of the bracing members and their connections, based upon the limit states of tension rupture on the effective net section and block shear rupture strength, as specified in LRFD Specification Section J4, shall be at least equal to the Required Strength of the brace as determined in Section 13.3a.

The tensile rupture limit state is given in Sect. J4 as φF_uA_e with φ = 0.75 (LRFD) and 0.5F_uA_e (ASD). For all standard steel grades, R_yF_yA_g is greater than the tension rupture strength. Does that mean that all bracing members need to be reinforced at the connection? If so, how far from the connection should the reinforcing extend?

I am analyzing and retrofitting a building from 1965. It is a one-story roof structure. Column bays are typically 30 ft by 60 ft. Steel bar joists are spaced at 3 ft o.c. and span the 30 ft dimension. Beams are cantilevered 9 ft over the columns with 42 ft infill beams spliced between the major girders. There is no bottom flange bracing or full-depth stiffeners at any point along the major girders. Are there any references from this era that describe this system and the design philosophy at the time?

We are reviewing the roof of a steel-framed building built in 1951, but we only have a few sheets of the original contract documents. The columns are referred to as 5-in. "Lally Columns." The column-beam connection is a single 1/4-in. steel knife plate through the column with six 3/4-in. bolts.

Was "Lally" used as a generic term for pipe columns? The connection described wouldn't work with today's Lally columns. It would seem to me that we are closer to a standard pipe column.

Table J3.3 in the 2005 Specification for Structural Steel Buildings gives dimensions for slotted holes. What is the definition for the slotted hole length? Is it the total length of the hole (i.e., from the outermost radius points on the circular portion) or the length of the straight portion of the hole?

Does the 2005 AISC specification cover the case of a WT member loaded in the plane of symmetry with the stem in flexural compression? Section F9 does not distinguish whether or not the stem is in compression or tension, but it does not seem to check the buckling of the web, which I would suspect is necessary for this case.

Is there a problem with hot-dip galvanizing high-strength structural steels?

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I am analyzing a building constructed in 1956. The plans specify high-strength bolts for the lift-slab columns, but don't give an ASTM designation. Do the A325 and A490 designations go back to 1956? Do you have any other suggestions for approximating design strength of these bolts without testing?

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I am currently working on modifications to an existing ship unloader at a terminal in the U.S., built around 1980. The design was done by a German company that has since gone out of business. The task at hand is to determine a demolition plan for part of the structure so that modifications can be made. The existing mechanical and structural drawings are all in German, but conversion of dimensions and loads are easy. However, many of the structural shapes are designated in a form that is not recognizable to us. We are accustomed to using metric equivalents used in modern construction; however, these members are designated in a non-standardized metric system notation.

For example, some wide-flange sections are noted as HEA 200; it seems obvious that this represents a 200-mm (about 8-in.) beam. Scaling the drawings confirms the depth, assuming they are properly to scale. Scaling flange width and flange thickness, it appears that this shape is nearly equivalent to a W8x35. I'm a bit nervous relying on scaling to confirm member sizes for strength and weight. There are other members designated HEB 200, HEA 500, etc. Is AISC aware of a steel reference that indicates these member sizes and section properties?

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This question originally appeared in the October 2006 Steel Interchange. AISC received follow-up correspondence from a fabricator to supplement the original response.

What is the recommended width tolerance of a slot in a tube shape that is to receive a plate? 1/8 in. larger? 1/16 in. larger?

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We specified HP12x53 for a pile foundation system. However, the contractor wishes to use W12x53 instead of HP12x53 due to the limited availability of HP12 shapes. Is there any technical reason not to permit the W12 substitution?

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I was recently performing fabrication inspection on a project, and the fabricator informed me that Type X bolts are slip-critical and Type N bolts are not. Is this true in general?

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What is a steel service center? Where can I find the names and phone numbers to some of the closest steel service centers to a particular location?

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Could you please advise on any special requirements for welding carbon steel to stainless steel? I have a condition where I have stainless steel plates that are to be welded to the end of HSS carbon steel sections, and want to make sure this is possible. Are there any references available that you could recommend on this topic?

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I am designing a Special Moment Frame (SMF) in accordance with FEMA 350 and using a reduced beam section. In the past, and in the FEMA 350 document, the columns for SMF have been limited to W14s. However in a recent AISC seminar, the speaker indicated that it is acceptable to use deeper column sections (W24s, W27s, etc.) for SMF using the reduced beam flange connection (dog-bone). I could not find any documentation supporting this unless specific testing is required. Can you point me in the correct direction?

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In the third edition of the LRFD Manual of Steel Construction, on page 16.1-143, there is a table for design stress for compression members in relation to Kl/r. Where is this table in the Steel Construction Manual, 13th edition?

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Section J1.7 in the AISC specification covers "minimum strength of connections" and requires a factored load not less than 10 kips for connection design. Where does the "10 kips" come from? Is that empirical? Can I use this load for the connections of precast concrete structures?

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Does AISC provide a spreadsheet that calculates the section modulus of two channels welded back to back?

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I received an RFI requesting a modification that includes thermal cutting (torching) of the deck closure angles (diaphragm chords). The special inspections agency is requesting that we verify this modification. I am trying to research the effects of thermal cutting. Do you have any recommendations on where to look for information as far as codes or technical references?

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Current details show an all-around fillet weld symbol at wide-flange column to base plate connections. I thought that this wasn't a good practice. Could you please comment?

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Full penetration welded column splices of columns with 65 ksi yield strength were completed using 70 ksi weld metal, when 80 ksi weld metal was required. We realize something needs to be done as the weld no longer develops the strength of the base material. Can the columns simply be reinforced at the connection, i.e. reinforcing flange and web plates to resist for the excess not developed by the weld; or does the 70 ksi weld no longer have any analytical capacity because it used with 65 ksi material?

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I noticed that the workable gage for W8x24 and W8x28 in the 13th edition Manual changed from previous editions of the steel manual. The 13th edition shows the workable gage as 4". for those members; previous editions list them as 3 1/2". Is this intended or was it a typo? Also, I was wondering if you can explain the difference between workable gage and useable gage as was listed in older U.S. Steel or ASD manuals.

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Do you know of any resources or design aids for the design of solid plate floor decking such as "diamond plate"? We tend to do quick designs using ASD and 0.75F_y for bending; then check deflections - but this seems too conservative.

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I understand that changes have been made in AISC's Steel Construction Manual, whereby hooked anchor rods are to be replaced by headed anchor rods in tension applications. When does this become effective in the various building codes? What are the dimensional specifications for the head configuration? Has AISC published guidelines for material specifications, etc.?

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The table User Note F1.1 lists yielding, flange local buckling, and web local buckling as the limit states for rectangular HSS in bending. Are rectangular HSS in flexure not affected by unbraced length? Is there no lateral-torsional buckling mode? I know that HSS are good for torsion, but as you get to some oblong sections (say an HSS 16x4), I would think that lateral-torsional buckling should begin to affect the strength.

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Why does Table 7-2 in the 13th Edition Manual use the nominal bolt area and not the net tensile area in determining the available tensile strength of the bolts? If I were to use a 1" diameter ASTM A490 bolt, Table 7-2 indicates a nominal area of 0.785 sq.inch and a tensile strength of 66.6 kips. But, if calculated based on the net tensile area of 0.606 sq.inch given in Table 7-18, the strength would be 51.4 kips. What am I not seeing?

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We seem to have a recurring issue with camber in composite floor systems. We mechanically camber the member in the shop, and our QC staff visually verify with a string line and tape that we have induced the specified camber per the shop/contract drawings. However in a number of cases, during surveys of the in-place members, as much as half of the camber we originally induced has relaxed out of the member.

We are familiar with Section 6.4.4 and the Commentary in the AISC Code of Standard Practice, and frequently provide our clients with a copy of it. We frequently ask the client to survey the floor prior to placing that concrete to address any related issues so that we can shore beams, if necessary. As indicated above, that's when we find that the camber is no longer the same as it was when it left the shop.

We're looking for any type of information that would help support the commentary in section 6.4.4 that it can only be measured in the shop and how to overcome the end user's perception of expected floor performance.

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Is it correct and permissible to use the "actual" tested yield strength of a square HSS (e.g., F_y = 49.3 ksi) in lieu of the "design" yield strength of ASTM A500 grade B (46 ksi) for the calculation of the flexural strength?

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Is the minimum bolt edge distance on a steel base plate 2d (where d is the bolt diameter)? Is this typical? What about the relationship with the base plate thickness?

I am reviewing a set of existing drawings dated 1953 to investigate proposed structural framing modifications. The drawings reference a B-3 connection. Is there a detail available of this connection?

When you are using snug-tightened bolts, is the proper installation method the "turn-of-the-nut" method?

I was just looking up some preliminary column sizes in the 13th Edition of the AISC Manual and the W tables have been truncated at W8x31. I could have sworn there were smaller sizes, so I looked at the 2nd Edition LRFD Manual; it includes W8x24 and also lists W6s, W5s, and W4s. Have these shapes largely been supplanted by HSS sections and therefore deemed not worthy of book space? Why the change?

I have an elevated walkway (20 ft high) that is foundation-supported. I understand the building code considers this a non-building structure. IBC 2003 refers you to ASCE 7 section 9.14.7, which gives you a different Table (9.14.5.1.1) for R values. I am using braced frames as the lateral system. For concentric braced frames ASCE refers you back to the typical Table 9.5.2.2. Do I now select special/ordinary/etc. braced frames and comply with all respective detailing, thus treating the walkway just like a building?

Equation K1-10 (2005 Spec) appears to be the only check required on the HSS wall for shear tabs on rectangular HSS shapes. Can you confirm that bending of the HSS wall does not need to be considered?

Also, for pipes or round HSS, what are the limit states for checking the HSS for a shear tab connection? For smaller pipes, it seems that arching will help, but as the diameter increases, and the surface becomes flatter, the shape more closely resembles a flat surface, like a rectangular HSS.

What is the recommended width tolerance of a slot in a tube structure that is to receive a plate? 1/8 in. larger? 1/16 in. larger?

During the design phase of a project our office is doing, we were asked by the client representative if choosing steel as the construction material (in lieu of concrete) will cause any potential wireless interference (the project is a hospital). The other concern was possible audible vibration (due to human activities and mechanical systems).

Are there any criteria in AISC 341-02, Seismic Provisions for Structural Steel Buildings, that would prohibit a beam splice outside the protected zone for an IMF in SDC D?

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1. Is there any difference in hole diameter for an expansion bolt vs. a normal bolt?
2. Is there any difference in hole diameter for a tension control bolt (TC), vs. a normal bolt, vs. a slip critical bolt (SC)?

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Is there a minimum strength in tension required between a column and a footing even though the column will never see any tension?

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I am working on modifications to an existing hotel in Manhattan that was built about 1928. I have access to some of the erection plans which indicate beam sizes for which I have no reference data, including a "CB" designation. Where can I find information on old rolled shapes?

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What ASTM specification should I use for steel pipe? Should I use A53 or A500? I previously used A53, but I read in an article in MSC that we should no longer designate pipe as, say, 4 in. diameter schedule 40, but rather as HSS 4.5×0.237, and that I should specify A500 instead of A53. Which is correct?

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What are the preheat requirements for fillet welding a 3.5"-thick base plate to a W14x283 column?

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I am trying to determine the load-bearing capacity of a roof on a building that is about 100 years old. The steel has been identified as S9x19.75 (purlins) and S15x33 (girders). Is there any way, for purposes of calculations, to determine the yield strength of the members? I'm guessing it's unlikely that the members are ASTM A36 steel. What was standard for the time?

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I recently received an RFI stating that one of the four 11/4 in. diameter ASTM A307 anchor rods at one braced column was bent out-of-plumb by 22° and asking for a fix solution.

In the past I have seen steel workers swinging big sledgehammers to straighten crooked anchor rods. However, I am hesitant to recommend this practice. I would like to recommend that they heat the offending rod and bend it gently back into place using a large piece of pipe as a lever. Is this an acceptable way to straighten slightly bent anchor rods or is there a preferred or published methodology? Some related questions are "If heat is used, how much should they heat the rod?" and "Could this procedure be utilized for various grades of anchor rods?"

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Are angle shapes produced in Grade 50 material?

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I would like to know the industry standard conversion for 1" ASTM A325 Imperial bolt to a Metric bolt. Please indicate the standard metric size bolt.

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I would like some clarification on the seismic design provisions of AISC 341-02, AISC LRFD-99 and the provisions of IBC 2003. Table 1617.6.2 of IBC permits designing steel structures as "Structural Steel Systems not Specifically Designed for Seismic Resistance." IBC 2205.3, "Seismic requirements for composite construction," states that in Seismic Design Category B or above, the design of composite systems shall conform to AISC 341, Part II. If I'm in a SDC C, and want to avoid "detailing for seismic", if I use R = 3, Ω₀ = 3 and C_d = 3, can I design a composite system per AISC LRFD-99, without following AISC 341 Part II?

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I have a building in which I used X-braces to transfer the lateral loads to the foundations. In a few bays, I have to move the bottom of the braces up three feet from the finish floor elevation to allow access for doors. This building is in a high seismic area (Seismic Design Category E), and is a one-story building (approx. 18 ft to bottom of steel). Can this still be considered an Ordinary Concentrically Braced Frame?

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Table 10-9a of the 13th edition Manual appears to incorporate a vertical edge distance of 1 1/4 in. instead of 1 1/2 in.

I had read that the hole in shear tabs can get rounded up to 1/4 in. when the bolt goes into bearing. Also, when the plate is sheared, there is 1/4 in. of material that may be "mushed," and not counted as part of the material, which is the reason why you are permitted an edge distance of 1 in. for flame or saw cut edges, as compared to 1 1/4 in. for sheared edges. If this is true, then the final calculable vertical edge distance for the bottom bolt on these plates will be 3/4 in., after removing 1/4 in. for the "mush" from the shear, and then 1/4 in. for the vertical elongation from the bolt in bearing.

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We have a large manufacturing building that was designed in 1945 and built in 1946. Is it true that A7 steel with a yield stress of 33 ksi was used at this time and until 1964? Is it acceptable to use current design methods for allowable bending, tension, and compression using 33 ksi for the steel yield strength? The "Historical Record, Dimensions and Properties-Rolled Shapes" published by AISC and edited by Herbert Ferris references the 1936 AISC specification and states the allowable basic working stress to be 20 ksi. This appears to come from 0.6 F_y (assuming 33 ksi steel). Does this mean F_b = 0.66 F_y could not be used if applicable? We have a copy of the AISC fifth edition specification published in 1947. Were the unit stresses shown in Part III of the specification the same in the fourth edition, which we are assuming was used until 1947?

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While there may be some structural advantages with HSS, there seems to be growing concern in the industry with fireproofing HSS when they are used as structural beams. Is spray-applied fireproofing a suitable method for protecting HSS beams? There do not appear to be any UL designs for fire protecting HSS beams. Can you address this common issue?

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Table J2.4 in the ninth edition ASD manual and the third edition LRFD manual shows the minimum sizes of fillet welds based on "thicker part joined." The new 13th edition manual is now based on "thinner part joined." Why has this changed?

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In the ninth edition ASD manual there is a table on page 1-8, Table 2. In the LRFD third edition, the table is found on page 2-27, Table 2-4. Can you please tell me where this table is found in the new 13th edition AISC manual?

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Is it now permissible to use plasma or flame cutting methods to make bolt holes?

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We are designing a structure in accordance with IBC 2003, which references the AISC 341-02 Seismic Provisions for Structural Steel Buildings. We have an ordinary steel concentrically braced frame and are in Seismic Design Category C. Per IBC 2003 Table 1617.6.2, for an ordinary steel concentrically braced frame, special detailing should be provided per Chapter 14 of AISC 341-02. This chapter requires connections of bracing members to be connected for an expected tensile strength of R_yF_yA_g. However, in Chapter 1 of AISC 341-02 under "Scope," it states that "These provisions shall apply to buildings that are classified in the applicable building code as Seismic Design Category D (or equivalent) and higher or when required by the engineer of record." Do we need to detail according to the Seismic Provisions requirements in Chapter 14 per IBC, or does the statement in Chapter 1 of AISC 341-02 allow us to not follow these provisions because we are in Seismic Design Category C?

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The AISC manual lists paper as weighing 58 lb/ft³. I have spoken with people in the file storage industry (condensed filing) and they use 40 to 43 lb/ft³. Do you know where AISC got the value they use and what it represents?

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How do you design anchor rods for combined loads of tension, shear, and bending? What equation do you use to combine the forces?

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Is there any concern relative to using a cruciform column (symmetric column in both directions with one column web cut and welded to the other column web) in an SMF? If so, are there any design criteria or is there research regarding the use of this type of column? The column and beam combination in either direction are pre-qualified per FEMA 350.

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I have a building that was designed as a special concentrically braced frame (R = 6). It is a one-story building, and most of the vertical braces were detailed to be shop or field welded. However, there are some horizontal connections that are to be field bolted. The AISC Seismic Provisions for Structural Steel Buildings require that the surfaces be prepared for slip resistance (Section 7.2), which we did. However, the contractor has accidentally painted over these holes, and the building is now erected. Would it be possible to use primer paint in place of a reduced coefficient of friction (lower than 0.33) to justify that the bolts have a lower than class A coating, but still satisfy the need to be slip-critical for the building?

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I understand the capacity of a threaded rod in tension is equal to the gross (nominal) area times 0.33F_u. Does the 0.33 factor include the effects of the potential for stripping of the threads of the rod through the nut placed at the rod end?

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I just received my copy of the 13th Edition Steel Construction Manual. Am I misreading these tables or have all of the connection capacities for ASD connections really been beefed up this much? For example, a shear tab connections with a 3⁄8 plate with two 3/4 bolts now can carry 21.2 kips.

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Section J1.5 of the 9th edition ASD Manual of Steel Construction states: "The connections at ends of . . . members in trusses shall develop the force . . . not less than 50% of the effective strength of the member . . ." Chapter 13 of the 13th edition Steel Construction Manual seems to indicate this is no longer required (see the paragraph titled "Minimum Connection Strength"). Is this correct?

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A contractor on one of my projects has substituted ASTM F1554 threaded rod with loose nuts in lieu of the specified headed anchor rods. Some of the rods are resisting tensile loads. It is my understanding that the nuts were not welded to the threaded rod as recommended by AISC Design Guide 1: Column Base Plates. Are there any additional recommendations for corrective action for anchor rods with loose nuts embedded in concrete that will be used to support tensile loads?

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A response given for the question "Axial Compression Capacity" in the March 2006 edition of Steel Interchange (available at www.aisc.org/steelinterchange) was, in part, as follows: "For the 2005 AISC specification, we recognized that UM plates are no longer available and eliminated them from the determination of the resistance factor and factor of safety. As a result, phi is 0.9 and omega is 1.67 in the 2005 AISC specification; hence, the difference in strength you noted." I have checked both LRFD and ASD versions of the steel manuals cited below and here are the results:

The explanation of the increased phi value for the LRFD version seems okay, as the new 13th edition manual states a higher load capacity compared to the previous LRFD manual. However, as you can see, the new ASD table in the 13th edition appears to have a lower capacity value compared to the previous ASD manual, which is the 9th edition. A decrease in omega, as explained in the article, should have resulted in a higher capacity in the new ASD table.

What explains the decrease in load capacity, comparing the new 13th edition ASD table with the old 9th edition for the member size W14×132 and effective length of 30'?

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On page 10-124 of the 13th edition manual, r_n = the nominal strength of one bolt in shear or bearing, kips. Please define "or bearing." Which bolt edge distance governs: the bolt closest to the edge of the connection material or the bolt in the interior of the connection? Furthermore, the specification on page 16.1-111 specifies, "For connections, the bearing resistance shall be taken as the sum of the bearing resistances of the individual bolts." Which is it?

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I am searching for a method to calculate the I_z of an L6×4×3⁄8. The r_z is listed in the LRFD manual; but not the I_z.

Where can I find historic information on old AISC manuals and standards like the AISC specification, Code of Standard Practice, RCSC specification, and common ASTM standards? Is there a separate list of the dates of publication of these standards?

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When applying the amplified seismic load with a system overstrength factor of 2 as required by Table I-4-1 of the 2002 Seismic Provisions, what overstrength factor should be used for the roof diaphragm? Do I need to apply the same factor?

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Where can I find the allowable shear stress values for ASTM F1554 grade 105 anchor rods?

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What effective length should be used for diagonal double angle cross bracing? At the intersection, one set of double angles is continuous and the other is bolted or welded to a gusset plate. I reviewed literature that indicated the tension diagonal can be used to brace the compression diagonal. However, this was for single-angle cross bracing where each member is continuous at the intersection. It would be conservative to just use the diagonal length for out of plane buckling, but that may be more conservative than necessary.

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What are the requirements for column-beam moment ratio for ordinary moment frames, if any, for Seismic Design Category A, B, or C? Do the strong column/weak beam requirements for special moment frames apply to OMF as well? Also, if there are requirements, do they change whether one is using the Seismic Provisions or the Specification for Structural Steel Buildings?

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Is there any concern relative to using a cruciform column (symmetric column in both directions with one column web cut and welded to other column web) in an SMRF? If so, are there any design criteria or is there any research regarding the use of this type of column? The column and beam combination in either direction are pre-qualified per FEMA 350.

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I am looking for design section properties for steel angles with respect to the principal major and minor axes (as opposed to the x and y axis listed in the Manual). Is there a reference for this?

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The AISC Seismic Provisions state that these provisions are applicable for structures in Seismic Design Category (SDC) D or higher. Does this mean that even if one uses R > 3 in SDC A, B, or C he or she does not need to apply these provisions? At the same time, the same paragraph ("Scope") says that the provisions are intended for structures for which the seismic design forces have been determined in assumption of inelastic response. This tells me that if one uses R = 5, thus relying on inelastic response, he or she should use the provisions regardless of the SDC.

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What is the formula for allowable stress for strong axis bending of a flat plate? I can find the weak axis of 0.75F_y, but no strong axis.

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What grade of steel rivets was used in 1925 steel construction? I am presently checking an existing steel girder's riveted connection capacity. I would like to confirm the allowable tensions (F_t) and allowable shear (F_v) of the 1925 rivets.

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The example column design problem from Charles Page's article in the November 2005 issue ("SpecWise: Design Examples," available at www.modernsteel.com) indicated that the 13th Edition Manual of Steel Construction lists an LRFD axial compression capacity of 892 kips for the W-shape. Looking up the same column shape and effective length in the LRFD third edition manual, I found that the compression capacity is 844 kips. This is a +6% of capacity. I did not know that increased capacity was incident to this specification change. Is that the case?

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What is the rule of thumb for how far bolts should extend above the nut--two threads, 1/2" to 1"? I saw in an old structural steel detailing book from the 1970s that this should be around 1/2" to 1". In looking at newer publications, I have not found any recommendation on how far these should protrude.

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Why does the ASD ninth edition manual require the use of C_b = 1 when computing F_bx to be used in equation (H1-1) for frames braced against joint translation? Generally, the columns in frames braced against joint translation will have bending moments in-span smaller than bending moments at the ends. On the surface it would seem to be an appropriate application of the C_b term as a reflection of moment gradient.

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Is it preferable to provide short-slotted holes on at least one end of cambered floor beams? I have been told that this practice allows subtle beam end rotation to take place as the concrete is placed and the camber is relieved.

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We use OMFs with bolted end-plate moment connections (neither the flanges nor the web of the beam are directly welded to the column). Section 11.2c of the 2005 Seismic Provisions lists three specific cases when the CJP groove weld is demand critical. However, none refers to end-plate connections. This is also consistent with the user note in 7.3b. I would appreciate further explanation, as it appears that the 2005 Seismic Provisions impose no additional welding requirement for the type of end-plate moment connections we use in OMFs.

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Are hollow structural sections available in F_y = 50 ksi steel strength? If yes, how do you specify the HSS?

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Does AISC have a test procedure for performing a load test on in-place steel? If yes, what publication is it in?

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I am designing a rigid-frame structure in California. I have heard that AISC has published the 2005 version of the Seismic Provisions and a new Seismic Design Manual with a list of prequalified connections. Is it true that these new connections will allow columns to a greater depth than the 14" limited by FEMA 350?

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Why does Appendix 6.3.1 of the 2005 AISC specification mandate that bracing be attached near the top (tension) flange of cantilevered members?

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I have structural steel that will receive spray-applied fireproofing. To what specification should the steel be cleaned, if at all?

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We are working on a waste water treatment plant. The structural drawings show pipe supports that are fabricated from structural shapes. They also show a detail referencing the mechanical drawings. The original bid set of drawings called for this material to be mild steel, but there was an addendum issued during the bid process. I reviewed the addendum for architectural and structural references, drawing numbers, or specification sections and found nothing referencing the pipe supports. This information was sent to our detailers and they also reviewed the addendum for architectural and structural references and found nothing. An addendum to the mechanical drawings referenced changes to this material. The material was revised from mild steel to stainless steel. Is it common practice for the fabricator to check all drawings and spec sections of the addendum for work related to architectural and structural steel?

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Figure C-I-13.2 of the 2005 AISC Seismic Provisions shows the brace end offset from the gusset bend line. This is recommended for SCBF. Would the same apply to OCBF or is there a different detail?

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When using LRFD, is there a specific interaction relationship between loads applied to a weld both parallel and perpendicular to the weld at the same time? Can the full calculated capacity be taken in each direction, or must the loads be combined?

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What is the correct method for repairing a bent 1" diameter A36 anchor rod? The anchor rod is one of a set of four for a W10 column in the final structure. The bend is approximately 60 degrees and is located in the threaded region of the bolt, with no apparent cracking or kinks at the bend. Heating and bending has been brought up, but not decided on. Is this an adequate solution? If so, what codes or guidelines are available that discuss temperature, bending procedure, etc.

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My customer requires me to supply Charpy test results on raw material steel. Where can I get this information? The information is not available on the mill tests.

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Are there any provisions for flame-cutting holes into steel members regarding residual stresses?

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I am trying to find the torsion in an HSS 12×12. I am having difficulty finding a value for the modulus of ridgity. I checked your manual and I couldn't find a value. Is this information available? The steel is governed under ASTM A500 grade B.

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I have a column (gravity loads only) where the projected length of the anchor rod is too short. At best, the top of the nut is level with the top of the bolt. However, on some, the nut is 1/8"; above the top of the rod. Is there a way to determine the capacity with a partially threaded nut?

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How are slotted holes applied -- to the centers of the round portions of the holes (the dimensions given in the AISC Specification) or to the edges of the slot (out to out)?

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All steel books I have read list lengths of slotted holes as either short or long. How about the lengths in between? The shop where I previously worked used 13/16" × 1 1/16" slot punches. Would this still be considered a short-slotted hole? I have always asked, "When does the length change from short to long?" No engineer whom I have asked has been able to provide an answer. How about slots longer than 13/16" × 1 7/8"? I would assume that it would be treated as a long slot. Is that right?

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When welding flat surfaces to rectangular HSS, a flare-bevel situation is created. The effective throat size is defined by Table J2.2 as 5/16 of the radius. I found a little information on this radius in the appendix of the HSS specification, but is there a comprehensive table somewhere? It seems like the inner and outer radii should be included in Table 1-11 with the rest of the dimensions.

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I am interested in using concrete filled HSS, but I am concerned about load transfer between the steel and concrete. Specification I2.4 addresses concrete encased columns but is silent on concrete filled HSS. Commentary I2.4 states that bond is commonly used on fixed offshore platforms, but no guidelines are available for other structures.

My application is "other structures," and in my application the load is applied to the HSS. Would shear connectors be required to ensure composite action, or can bond be used on other structures? What would be used for shear connectors on HSS, and what are the design criteria? How is bond stress evaluated?

Question sent to AISC's Steel Solutions Center

I attended an AISC staggered truss seminar, and I have questions regarding some braced frame requirements: We are doing an industrial structure in California, Zone 3. We cannot find anywhere in the 1997 UBC, 2001 CBC, or AISC seismic provisions whether we can use an OCBF in a Zone 3, or if the lateral system must be a SCBF. There is not a lot of difference between the two, but we have approximately six braced bays on six different levels. To be able to use our OCBF spreadsheet would be helpful.

Question sent to AISC's Steel Solutions Center

Is there any new thinking or literature regarding anchor rods loaded primarily in shear? This is especially important in metal buildings where a shear key welded to the base plate normally does not exist. How do you handle the effect of oversized holes in base plates? I normally ignore this (unless the loads are high) and assume the shear forces are essentially resisted by the anchor rods.

I can come up with mechanical or welded connectors of some type, but it seems to be a needless expense, given that failure of anchor rods in shear does not seem to be a recurring problem. I was told that there are some new guidelines from AISC, but I haven't found them on the web site.

Question sent to AISC's Steel Solutions Center

The Canadian Institute of Steel Construction recently presented a seminar on steel plate shear walls at our office. They mentioned that there is an AISC committee studying this subject. Is there any information on design procedures and R values for seismic load determination in accordance with IBC 2000 or IBC 2003?

Question sent to AISC's Steel Solutions Center

I am curious about the origin/calculation behind the equation for a trial beam size shown on page 5-26 of the AISC LRFD manual.

Question sent to AISC's Steel Solutions Center

We are designing the rehab of several old buildings. In at least two of them, we have structural steel framing that we want to weld to. One was built in about 1901, the other in 1912.

We are in the agonizingly slow process of getting the CM to contract with a testing firm to test the steel for weldability. The CM is not overly concerned because he says that almost any steel is weldable -- it's just a matter of picking the right electrode. Sounds too good to be true. Any comments?

Question sent to AISC's Steel Solutions Center

When using a field bolted/field welded moment connection, is it AISC standard practice to leave the backing bar in place after the weld is complete?

Question sent to AISC's Steel Solutions Center

Are there any standards for the minimum edge distance required for column base plate holes?

Question sent to AISC's Steel Solutions Center

I have a column with a hot-rolled crane bracket fillet welded to the inside column flange. Table A-K4.2, Case 21, in the 1989 ASD specification seems to apply. I used the elastic (vector) approach to design the web-to-column flange weld. Should I just take the applied kips/in. that I used for the welds and divide that by the web thickness to get a stress in ksi? That seems like a slightly simplified and conservative method.

Question sent to AISC's Steel Solutions Center

I am being asked to help resolve a dispute between a steel erector and another subcontractor who is installing handrail to the side of a vertical C channel. I believe that the top of this channel got pulled in during erection so the web is no longer exactly vertical. I can't find anything specifically addressing this case in the Code of Standard Practice for Steel Buildings and Bridges (COSP), but I am probably just missing something.

Question sent to AISC's Steel Solutions Center

The definition of the seismic load resisting system (SLRS) in the 2002 seismic provisions glossary includes collectors. Does it follow that Section 7.2 (seismic bolting requirements) applies to collector design? If yes, then how can one design an axial bolted connection for a ductile limit state?

Question sent to AISC's Steel Solutions Center

I understand the detailing and fabrication requirements for weld access holes in moment connected beams have changed in the last few years. Could you give me information on what the new requirements are?

Question sent to AISC's Steel Solutions Center

Is there data providing support to the theory that welding to a sheared edge of a connection angle is acceptable? Angle thickness will vary from 5/16" to 3/4".

Question sent to AISC's Steel Solutions Center

I am working on a remodeling project in an existing building that was built in the 1905 to 1915 time range. I field measured an I-shaped beam as follows: d = 15", bf = 5.5" and tf = 0.5". Do you know its properties and designation? Its Fy?

Question sent to AISC's Steel Solutions Center

I'm looking for some documentation that states whether the contract drawings or the project manual (or specification) dictates when there's a conflict. The architectural drawings call for painted lintels, but the spec calls for galvanized lintels. Which one takes precedence?

Question sent to AISC's Steel Solutions Center

We are currently detailing our first Special Concentric Braced Frame (SCBF) in Seismic Design Category D and have a question. We understand the concept of creating a yield line through the gusset plate. Our issue concerns the column base at the floor slab. The gusset plate will need to be huge to establish the yield line above the concrete slab because the slab would confine the brace and gusset, preventing the yield zone from forming. The best idea we can come up with is to wrap the gusset and brace end with a layer of compressible material, such as rigid insulation, prior to pouring concrete. Is this something that has been discussed or written about already? Does our idea seem reasonable? We are assuming that slab confinement is an issue here and would like to know if it is typically dealt with or simply ignored.

Question sent to AISC's Steel Solutions Center

I am currently working on a project where an installed anchor rod was bent during backfilling against a concrete wall. The anchor rod projection from the concrete was bent to a 45 degree angle and the contractor would like to "slowly heat the rod and straighten it." Is this an acceptable repair? It seems that this may weaken the rod. Is it better to remove the anchor rod in its entirety from the concrete? What is the typical repair for this type of damage?

Question sent to AISC's Steel Solutions Center

Is there an AISC requirement that structural steel shop and erection drawings be prepared under the supervision of and stamped by a licensed professional engineer in the applicable jurisdiction, or that the drawings are to be reviewed and stamped by a P.E.? Is either of these criteria a requirement of AISC?

Question sent to AISC's Steel Solutions Center

I thought the LRFD values for bending should be FyZx rather than FySx as shown in the web site document "Basic Design Value" (available at

Question sent to AISC's Steel Solutions Center

I have a situation with an HSS 10x6x3/8 fitted with WT connections that enclose both ends. It is exposed to the weather so it has been specified as hot-dip galvanized. The question has been asked if the HSS will need vent holes, and, if so, how many and on which side for dipping? I am of the opinion that we will not need holes. Am I correct in this assumption?

Question sent to AISC's Steel Solutions Center

Is it permitted to run electrical conduit in the concrete floor slabs of composite steel beams? If so, how does one determine or specify an amount or size of conduit permitted? What if a large amount of conduit crosses perpendicular to a beam at either the center span, where compression is a concern, or the near the support, where shear transfer isa concern?

Question sent to AISC's Steel Solutions Center

There was a Steel Quiz answer (July 2005, question nine) that stated seismic moment end-plate connections are to be prepared as slip-critical bolted joints, but designed as bearing joints. The answer refers to Section 7.2 of the 2002 AISC Seismic Provisions for Structural Steel Buildings, but not the corresponding Commentary, which better explains the provision and specifically relaxes the requirement of faying surface preparation for moment end-plate connections.

Question sent to AISC's Steel Solutions Center

I am working on a project that was built in 1957-1958 and I am trying to check the existing column anchor bolts for shear. I have the 6th Edition Manual of Steel Construction (1967) which gives shear values for ASTM A307, A7, and A373 steel (Fv = 10 ksi). Do you know what type of steel (Fv) was predominately used in the late 1950s for anchor bolts?

Question sent to AISC's Steel Solutions Center

My firm specializes in inspection and materials testing. I am looking for a publication that covers the basics of structural steel inspection. Do you have any recommendations from AISC's publications or other sources?

Question sent to AISC's Steel Solutions Center

I am analyzing an existing steel truss constructed of various sizes of W14 beams. A few of the top chord members are over-stressed. Could plates be added between the flanges of the existing section to increase the effective ry? This would lower the kl/r value in the weak-axis and increase the capacity of the member. In a recent Modern Steel Construction, the issue of continuous welding vs. intermittent welding was addressed. Could intermittent welding be used in this case, or would a continuous weld be required?

Question sent to AISC's Steel Solutions Center

In Section J3.5 of the AISC 2005 Specification (www.aisc.org/2005spec), the maximum distance from the center of any bolt to the nearest edge of parts in contact shall be 12 times the thickness of the connected part under consideration, but shall not exceed 6". What is the basis of this requirement?

Question sent to AISC's Steel Solutions Center

Can you please expand on what is meant by the statement found in AISC Seismic Provisions for Structural Steel Buildings dated May 21, 2002 in Section 6.2 regarding when Ry can be applied to Fy in the determination of the design strength. The statement reads as follows: When both the Required Strength and the Design Strength calculations are made for the same member or connecting elements, it is permitted to apply Ry to Fy in the determination of the Design Strength?

Question sent to AISC's Steel Solutions Center

If we weld a heavy cross-section beam to column using a partial-joint-penetration (PJP) groove weld, does the member still need to have the required CVN values in the "k" area or is the CVN testing from the flange specimen good enough?

Question sent to AISC's Steel Solutions Center

We have a situation where connection plates are required at beam to column web moment connections. The beam flanges are attached to the connection plates with complete-joint-penetration groove welds. The connection plates are the same thickness and are the same grade of steel as the beam flanges. The bottom of the lower connection plates were detailed to be flush with the bottom of the bottom flange of the beam. The beams were overrun at the mill. The as-built condition is that the bottom of the bottom flange is lower than the bottom of the lower connection plate by 1/4 to 3/8 inch. Is there a tolerance allowance for a condition such as this? Can a small eccentricity between the center of the flange and the center of the connection plate be allowed?

Question sent to AISC's Steel Solutions Center

In the 3rd edition LRFD Manual, there are two different k values listed for a given wide-flange shape. One is kdetailing and the other kdesign. Why is this and which one should I use?

Question sent to AISC's Steel Solutions Center

Is there a standard chart to indicate the average embedded length for bent rod anchor rods?

Question sent to AISC's Steel Solutions Center

An engineer has specified ASTM A325 hooked anchor rods for a project. Would it be acceptable to use ASTM F1554, Gr. 105 hooked anchor rods as a substitute?

Question sent to the AISC's Steel Solutions Center:

Please provide guidance on how to calculate the allowable bending strength for tee sections. We are assuming that when the flange is in compression, the member can be treated as an I-shaped beam. What about when the stem is in compression or when the beam is subjected to weak axis bending?

Question sent to AISC's Steel Solutions Center

The ACI -factor for LRFD bearing is 0.65 while AISC allows 0.60. Which factor is an engineer supposed to use?

Question sent to the AISC's Steel Solutions Center

We are currently detailing our first Special Concentric Braced Frame (SCBF) in Seismic Design Category D and have a question. We understand the concept of creating a yield line through the gusset plate. Our issue concerns the column base at the floor slab. The gusset plate will need to be huge to establish the yield line above the concrete slab because the slab would confine the brace and gusset, preventing the yield zone from forming. The best idea we can come up with is to wrap the gusset and brace end with a layer of compressible material such as rigid insulation prior to pouring concrete. Is this something that has been discussed or written about already? Does our idea seem reasonable? We are assuming that slab confinement is an issue here and would like to know if it is typically dealt with or simply ignored.

Question sent to AISC's Steel Solutions Center

I am going to use oversized holes in flange-plated moment connections to a column utilizing slip-critical bolts. I would like to allow the erectors some leeway by letting them use shim plates between the flange plates and beam flanges. My questions are:

1. Should shims be utilized at both the top flange and the bottom flange simultaneously, or just at either the top or bottom?
2. What is the suggested maximum shim thickness? Can I just use that shown in Figure 3-4c on page 3-8 of the ASD 9th Edition Connections Manual?
3. Is there a minimum shim thickness?
4. Based on that shown on page 3-12 of the Connections Manual, is it correct to assume that either conventional or finger shims are permitted, even if the bolts are slip-critical in oversized flange plate holes?

Question sent to AISC's Steel Solutions Center

We are designing a one-story strong beam/weak column moment frame. The small columns we specified are W6x25s. They work fine in terms of structural capacity, but the fabricator is having trouble making the connections. They requested that the design be changed to a strong column/weak beam moment frame. Any thoughts?

Question sent to AISC's Steel Solutions Center

For the case of steel intermediate moment frames in dual systems, are continuity plates required if the requirements of FEMA 350 equations are met,

or are continuity plates always required? When are stiffener plates not required in steel intermediate moment frames?

Question sent to AISC's Steel Solutions Center

I know that there is a different method to use in designing the base plates of an HSS column as opposed to that for a W-shape. What equation is used for HSS bases, and how does one determine the area of the plate needed?

Question sent to AISC's Steel Solutions Center

I have HSS16x16 concrete-filled tubes. In AISC's Design Guide 19, it appears that because the section is larger than 12", fire protection is required. Is there any additional research that includes larger sizes of HSS filled with concrete that will meet fire code criteria?

Question sent to AISC's Steel Solutions Center

According to Section I4 of the 1989 ASD Specification, the value of V'h for shear connectors used in composite construction is not permitted to be less than 1/4 the minimum of Equation (I 4-1) or (I 4-2). If the number of connectors provided is less than required to induce a horizontal shear capacity equal to 1/4 the shear capacity of the section, can V'h still be calculated and a composite section assumed or should the assumption be zero composite action with the steel beam resisting the total load?

Question sent to AISC's Steel Solutions Center

When designing a single-plate shear connection for a beam-to-HSS column (Figure 4-4 in the HSS Connections Manual), the limit state of punching shear in the HSS wall is checked. In the procedure for a single-plate connection to a girder web (Example 10.12 in the 3rd Ed. LRFD Manual), there is no punching shear check on the web. Are the restraints sufficiently different so it is not a likely failure mode? Has there been any research on the shear-plate connection to a web?

Question sent to AISC's Steel Solutions Center

Page 4-116 of the 9th edition ASD Manual indicates that an end plate moment connection can only be used for static loading. However, I have just purchased Design Guide 16, and I cannot find anything that would indicate that this connection must only be used for a static loading condition. Can this connection be utilized for seismic and wind moments?

Question sent to AISC's Steel Solutions Center

Can you guide me as to how to decide or choose between using continuous or intermittent fillet welds?

1. For a crane girder beam (I beam with channel on top) whether intermittent welding can be permitted.
2. For built up columns and beams whether intermittent welding can be considered and if so under what situations.

Question sent to AISC's Steel Solutions Center

We have been asked to supply an AISC erection plan on a U.S. Army Corps of Engineers project that requires a Certified Fabricator, but not a Certified Erector. Is there such a thing as an AISC Erection Plan?

Question sent to AISC's Steel Solutions Center

Pre-qualified moment connections in FEMA are for special and ordinary moment frames, not for intermediate moment frames. Must we use the pre-qualified FEMA special moment frame connection for an intermediate moment frame defined in AISC? Or can we use the FEMA pre-qualified moment connection of ordinary moment frame for intermediate moment frame defined in AISC?

Question sent to AISC's Steel Solutions Center

How will the new combined code handle the different loadings requirements for the two design methods? Surely the loading requirements for the LRFD method will not change.

Question sent to AISC's Steel Solutions Center

I am looking for the bevel angle, size, and length of weld, etc. for a moment connection.

Question sent to AISC's Steel Solutions Center

We are involved in the design of a roof framing system comprised of a two-way grid of trusses. The trusses have round HSS web members and round HSS bottom chords. There are other portions of the project that involve standard wide flange framing. In our specifications, is it sufficient to require that welders be certified per AWS D1.1? Is there any additional certification required for tubular members? We have a lot of critical welds connecting round HSS to other round HSS as a part of the roof truss framing, but I do not know if that alone requires additional certification.

Question sent to AISC's Steel Solutions Center

I am working on a project where the architect wants to specify Architecturally Exposed Structural Steel (AESS) for the braced frame. I am hoping there are some standardized requirements so that the architect, engineer, fabricator, and erector are on the same page. Does AISC have a specification for AESS? The best information that I have located (so far) is the May 2003 supplement to Modern Steel Construction.

Question sent to AISC's Steel Solutions Center

I seem to remember when an angle is welded to a gusset plate that the weld should be "balanced" about the angle's neutral axis. Can you direct me to a code reference or design guide that confirms my recollection?

Question sent to AISC's Steel Solutions Center

HP Shapes

What does a steel shape that is called HP 10x57 look like? We have never run across this before nor has our steel supplier.

Question sent to AISC's Steel Solutions Center

We have designed a moment-resisting frame consisting of wide flange beams and HSS columns. The applicable building code is IBC 2003. We have used an ordinary steel moment frame (OMF) with R = 3.5. We have detailed the connection to consist of column through-plates (horizontal) at the beam flange elevations, and complete penetration groove welds of the beam flanges to these thru-plates. The welds specified conform to AISC's Seismic Provisions for FR moment connections (including backing bar detailing, removal, and reinforcing welds). We received a phone call from a fabricator who told us he thought AISC does not permit using an R factor greater than 3 for moment frames with HSS columns. Is this true?

Question sent to AISC's Steel Solutions Center

This regards designing for block shear per page 4-8 of the 9th edition ASD Manual. Is it permissible to increase allowable loads resulting from these calculations by 1/3 for seismic loads per UBC 1612.3.2? Can these results be increased by 1.7 per UBC section 1633.2.6 for collector elements?

Question sent to AISC's Steel Solutions Center

Do all components require inspection prior to leaving the fabricator for erection, even if it has been verified that the fabricator meets all the requirements for certification?

Question sent to AISC's Steel Solutions Center

Does a welded stud take the place of a puddle weld on decking installation?

Question sent to AISC's Steel Solutions Center

ASTM A992 is noted in the AISC LRFD Manual as the preferred structural steel for W-shapes. Does A992 have the same welding properties as the old A36?

Question sent to AISC's Steel Solutions Center

The following is a brief paper presenting one engineer's opinion on design parameters for use with cantilever beams. The author has included a list of pertinent reference sources on the subject.

Cantilever Flexural Member Design
By Sam Eskildsen, P.E., Structural Design Group, Birmingham, AL

I am designing a steel structure with 960' x 450' plan dimensions. Where can I find the design guideline for setting and designing expansion joints? Is it possible to design the structure without expansion joints if the architect does not want the joints?

Question sent to AISC's Steel Solutions Center

I am looking for information on what the minimum flange thickness requirements are for welding shear studs for composite action. Because I do not know what the actual weld size is, I cannot determine what the minimum flange thickness should be. We do specify 1/4" minimum.

Question sent to AISC's Steel Solutions Center

On one of our projects, the EOR specified snug-tightened bolts for the majority of the connections. The erector chose to use twist-off bolts for the entire job. He snapped off all the splines without telling the EOR. As the inspector, should I approve them?

Question sent to AISC's Steel Solutions Center

How do you calculate the lower flange loading capacity of a steel beam to be used to support an underhung crane? Are there any published ASD or LRFD design procedures?

Question sent to AISC's Steel Solutions Center

When shear is considered in a direction perpendicular to a bolt holes edge distance can the edge distances of table J3.4 of Specification Chapter J be reduced? If so by how much can it be reduced and is there a reduction in load capacity.

Question sent to AISC's Steel Solutions Center

Preheat and Stud Welding (from November 2004)

I have two questions:

1. When welding headed stud to beams or plates using the standard welding gun, are there any requirements to preheat the base material in cases where the base material is, say, 2" to 3" thick or more? I know that when the stud is fillet welded, AWS specifies the base material to be preheated but there is no mention of preheat when the stud gun is used.
2. Lastly, what problems can be expected if proper preheat is not provided and conventional fillet welds are used? Cracks in base metal? Cracks in weld? Other?

Question sent to AISC's Steel Solutions Center

Rehabilitation of ASTM A9 Steel and Rivets (from September 2004) We are reinforcing connections of an existing building from 1925 and have a few questions regarding the design approach that should be taken:

1. We have done coupon tests for a beam and a column and the lab qualified the structural steel as ASTM A36 with Fy = 36 ksi. Can we assume that the angles and plates forming the different connections are made of the same material?
2. The majority of the connections are riveted. Others are bolted. Can we assume that the riveted connections are slip-critical and therefore can be combined with new weld to enhance the connection capacity? In this respect we assume that the bolts are in bearing and their capacity should be ignored when reinforcing the existing connection with weld. Please advise.
3. Based on an old AISC manual we have found that the maximum 3/4" diameter hand driven rivet capacity in shear is 4.42 kips. We assume this is a slip-critical value. Is there any corresponding bearing capacity or does it not exist in rivets.
4. The rivet capacity in shear based on the old AISC manual is controlled by bearing (subject to the supporting member thickness) indicating different values for single and double shear. Are those values governed by the bearing on the rivet hole or bearing on the rivet itself. We are uncertain since the rivet material is weaker than the structural steel.
5. We are considering reinforcing the riveted shear connections in two ways: a.Replace old rivets with new high strength bolts. b.Add weld around the connecting angles. If both ways are acceptable to the contractor, which solution is more cost effective?

Question sent to AISC's Steel Solutions Center

Regarding single-angle connections with the "girder" leg welded and the "beam" leg bolted, the strength is apparently based on eccentric shear in the plane of the weld. Why is the eccentricity normal to the plane of the weld (distance to the bolt line) not also considered?

Question sent to AISC's Steel Solutions Center

Anchor traditionally have holes in base plates and in the anchor chairs/seats larger than is customarily provided for steel-to-steel connection bolts. What are the guidelines for edge distances for these holes? I have some field-enlarged holes for 1 3/4" dia. bolts in an anchor seat/chair that are 3 1/2" in dia. What edge distance criteria should I use? We plan to use 1"x5"x5" plate washers under the nuts for these bolts.

Question sent to AISC's Steel Solutions Center

What is the correct method for field cutting new openings in the web of a steel beam? The beam is already erected and the opening locations are acceptable.

Question sent to AISC's Steel Solutions Center

For a building constructed in 1961, what would have most likely been the specified yield stress for the steel? Would steel meeting ASTM A7 have been the typical steel used? Would the yield for A7 have been 33,000 psi?

Question sent to AISC's Steel Solutions Center

I received a response from the AISC Steel Solution Center concerning the rotational ductility of an all-welded shear tab connection. The response stated that they did not recommend this detail due to its inability to provide adequate rotational ductility -- i.e. the lack of bolts eliminates plowing action against the tab holes as a means for accommodating beam end rotation. Another question has arisen, where the fabricator has detailed double angle connections that are welded to the beam web and then bolted (or welded) to columns or girder beams. The ninth edition ASD Manual, Chapter 4, clearly denotes the use of these two types of connections. Could you please explain why these two types of connections seem to be permissible when it appears as though the same situation as the "all welded shear tab connection" applies? Do these configurations assume the bolts in these connections can elongate to allow free rotation?

Question sent to AISC's Steel Solutions Center

I am reviewing a steel building constructed in 1950 (no existing drawings). I don't see any braced frames in the structure, but find that the beam-to-column detail is built as a standard double-angle three-bolt shear connection, also with a top and bottom angle with bolts to the column flange and the other leg welded to the beam flange. This appears to be a semi-rigid connection for a moment frame. Was this approach widely practiced back in the 1950s?

Question sent to AISC's Steel Solutions Center

I believe there is an inherent difference between the "classic" shear tab with one end welded and one end bolted, and an "all-welded" single plate connection, correct? Does a bolted/welded shear tab allow rotation via bolt hole deformation and/or bolt oversize? I haven't found any information on single plate connections with both plate ends welded. My concern is that these all-welded connections allow end rotation of the supported beam. I have a construction project where the detailer specifies single plate, all-welded, simple shear connections. I spoke with him and he told me he details this type of connection all the time. He insists this isn't a problem because the beam web (in this case 0.20" thick), will experience a small amount of local yielding before the welds or shear tab yield (plate is 5/8" thick), which will allow for a small end rotation and hence provide a "simple shear" connection. Is this argument valid? I'm concerned the web could rupture or yield excessively.

Question sent to AISC's Steel Solutions Center

I am looking for some insight into the tension shear interaction for ASTM F155-Grade 55 anchor rods. Using ASD, Table J3.2 indicates the allowable stress on fasteners and Table J3.3 indicates the allowable tension based on the shear, however these tables are specific to A307, A325, A490, and A449 bolts. Since the transition to the ASTM F1554 for anchor rods, is there an updated recommendation for this interaction? If there is not an updated recommendation, is the use of the interaction based on a corresponding material with a similar Fu, the most appropriate solution?

Question sent to AISC's Steel Solutions Center

I am designing a moment connection between a W24x117 beam and a W18x46 column. The W18x46 column was sized to fit in a 6" wall. The beam is in a duct area, so width was not an issue. The column requires web stiffeners and LRFD Specification Section K1.9 (1) requires that the stiffener width be equal to one third of the beam flange width plus one half of the column web thickness. This equates to 12.8/3 + 0.36/2 = 4.1", which is problem due to the fact that my column width is only 6" and the stiffener would be past the column flange. Is it permissible to put in the required width of column web stiffeners, say 4 1/4" wide, and then notch the web stiffener the thickness of the column flange so that it meets back up with the abutting flange of the connecting beam? I don't know if it is allowed to have the web stiffener stick out past the column flanges to meet the requirement of LRFD Specification Section K1.9 (1).

Question sent to AISC's Steel Solutions Center

I am trying to design a WT with its stem in flexural compression per the 1999 LRFD Specification. The member has a slender stem and I have calculated Qs in accordance with Appendix B5.3a. I have checked the flexural limit states of yielding and lateral-torsional buckling strength per Section F1.2c. Does the reduction factor Qs apply to the lateral-torsional buckling strength and the yield strength? Or does it apply only to the yield strength?

Question sent to AISC's Steel Solutions Center

Where can I find reference material (with paragraph number) to prove to the contractor that the galvanization layer from galvanized steel members must be removed before welding these members together? In other words, the presence of zinc in the weld may weaken the weld capacity and be a source of health concerns for the qualified welders.

Question sent to AISC's Steel Solutions Center

Table 4-13 (Composite HSS Compression Members) in the 3rd edition LRFD Manual seems to include HSS that will not satisfy the minimum thickness requirement found in Section I2.1 of the 1999 LRFD Specification. For example, for an HSS 16x16x3/8, the minimum t would be 0.368". This is greater than the listed tdesign = 0.349", but is less than the nominal t of 0.375". Should we be using the nominal thickness rather than the tdesign for the minimum t criteria (which would contradict what is currently used for W-shapes, for example) or should this HSS be "disqualified"?

Question sent to AISC's Steel Solutions Center

Is there a maximum thickness recommended for a grout pack (assume a non-shrink grout) under a column base plate? Assume there are leveling nuts under the plate. Assume a multi-story structure. Would the grout thickness be predicated on the thickness of the base plate and the open space beneath it? Assume the method or placement (packed, poured, injected) will adequately fill the space between the concrete pier, footer, etc. and the underside of the base plate.

Question sent to AISC's Steel Solutions Center

If you take away most of one side of a top flange of a steel beam, say an 8" flange and on one side you notch out 3", what are some acceptable methods of reinforcement? There will be a pipe going through the notch, and I cannot do anything from the top. Is welding a plate of sufficient area to the underneath of the opposite flange ok? I see how this will replace the section modulus and compressive block that I am losing, but I am concerned with how the internal stresses of the beam "jump" from one side to the other.

Question posted on SEAINT (www.seaint.org)

We are reviewing our detailing standards and, the question of what "exactly" is required by OSHA concerning double connections was asked. I am sure you have addressed this question many times and hopefully have an (end all) answer. Could you please help us to clear up the confusion we have concerning this issue?

Question sent to AISC's Steel Solutions Center

I have two questions:

1. When welding headed studs to beams or plates using the standard welding gun, are there any requirements to preheat the base material in cases where the base material is say 2- or 3-inches thick or more? I know that when the stud is fillet welded, AWS specifies that the base material must be preheated but there is no mention of preheat when the stud gun is used.
2. What problems can be expected if proper preheat is not provided and conventional fillet welds are used? Cracks in base metal? Cracks in weld? Other?

Question sent to AISC's Steel Solutions Center

I am required to use the 1997 AISC Seismic Provisions for a braced frame detail with typical loads and configuration, except that it has a significant axial transfer force from the beam to the column. The only way to get this transfer force from the beam into the column (sometimes to the flange and sometimes to stiffeners in the column web) is to weld the flange directly to the column. The weld is a CJP groove weld.

Is it still allowable to use pretensioned bearing bolts in the gusset to column connection (again, it might be to flange or web)? The gusset to column connection is either two angles or an end plate with shear plates along the column web.

My opinion is that pretensioned X-bolts are acceptable in this situation, since the weld is not sharing load in the same axis with the bolts and since the CJP groove weld in the beam flange to column, if properly detailed with proper weld access holes, will have sufficient ductility/flexibility to accommodate the small potential slip under full load. What about low-seismic situations?

Question sent to AISC's Steel Solutions Center

A composite steel beam (W12x19) with concrete on metal deck diaphragm 4" total thickness is loaded with uplift. Can we assume that although the slab is at the tension flange it still restrains the beam against lateral-torsional buckling, so that only local buckling of flange/web has to be checked?

Question sent to AISC's Steel Solutions Center

I have a project for which I am using flange-welded moment connections. The fabricator has asked us to look into the possibility of replacing the full-penetration weld from the plates to the column flanges with two-sided fillet welds, sized to achieve the same capacity as the full-penetration welds. However, my supervisor believes there is some issue with fillet welds in direct tension, and that he is not comfortable with this connection as proposed by the fabricator. Can you refer me to any publication that shows the fillet welds are not the preferred way of making the connection?

Question sent to AISC's Steel Solutions Center

Which is more economical to specify-shear tabs or through-plates when connecting a wide-flange beam to a HSS column?

Question sent to AISC's Steel Solutions Center

It is my understanding that the practice of placing knee braces from crane support columns to the underside of crane support rails is not acceptable (in the past this was done to reduce the unsupported length of the crane rail). Just what is the reason for elimination of the knee braces?

Question sent to AISC's Steel Solutions Center

I have a number of "cripple" beams where the engineer of record calls for the web and flanges to be complete penetration welded at the miter. Since the beam size is such that the flange and web are less than 5/16" thick, I would like to detail the piece with out prepping the flanges or web and use the designation B-L1b-GF. Is it reasonable to believe that most shops and the field can use this welding process?

Question sent to AISC's Steel Solutions Center

We have a project that is all braced frame construction. The anchor rods specified were ASTM A36 from 1 3/4 to 2 1/2 diameters. The design specifications are requiring us to pretensioned A36 anchor rods to the minimum yield point specified for A36 steel. They request that this be done by the turn of nut method as indicated by the RCSC Specification for Structural Joints using ASTM A325 or A490 Bolts. This does not appear correct to our people. We see no correlation between RCSC and A36 anchor rods. Please give us your thoughts.

Question sent to AISC's Steel Solutions Center

Section I3.5b in the 1999 LRFD Specification limits Nr to the "number of stud connectors in one rib at a beam intersection, not to exceed three in computations, although more than three studs may be installed." My question is does the above definition mean that more than three studs can be installed and counted on for strength but that only in equation I3-1 need the value for the parameter Nr be limited to a maximum value of three? Or does it mean that never more than three studs per rib at a beam intersection can be counted for any part of composite design?

Question sent to AISC's Steel Solutions Center

I am looking for the tensile capacity of an ASTM F1554 anchor rod grade 36. In LRFD, its nominal strength is listed in Table J3.2 as 0.75(0.75F)A. However it is noted at the bottom of the page that this is only for static loads. What value should be used if anchor rods are exposed to fatigue loads?

Question sent to AISC's Steel Solutions Center

Can I use A490 bolts in structures subjected to fatigue loads, such as bridges? I thought I remembered the answer was no.

Question sent to AISC's Steel Solutions Center

I am trying to clear up some confusion about fillet weld termination. We are fabricating beams for a 20-story building which have 1" gusset plates welded to top flanges of various large W shapes. The shop we are subletting some of the work to is terminating the fillet welds anywhere from 1/4" to 3/4" from the ends of the beams (welds vary from 5/16" in size to 5/8"). I say when a designer says to put fillet welds on both sides he means the total length of welds sides with no end termination. Could you please show me specifically where this issue is addressed and what exactly is acceptable?

Question sent to AISC's Steel Solutions Center

We are reinforcing connections of existing building from 1925 and have a few questions regarding the design approach that should be taken:

1. We have done coupon tests for a beam and a column and the lab qualified the structural steel as ASTM A36 with Fy = 36 ksi. Can we assume that the angles and plates forming the different connections are made of the same material?
2. The majority of the connections are riveted. Others are bolted. Can we assume that the riveted connections are slip-critical and therefore can be combined with new weld to enhance the connection capacity? In this respect we assume that the bolts are in bearing and their capacity should be ignored when reinforcing the existing connection with weld. Please advise.
3. Based on an old AISC manual we have found that the maximum 3/4" diameter hand driven rivet capacity in shear is 4.42 kips. We assume this is a slip-critical value. Is there any corresponding bearing capacity or does it not exist in rivets.
4. The rivet capacity in shear based on the old AISC manual is controlled by bearing (subject to the supporting member thickness) indicating different values for single and double shear. Are those values governed by the bearing on the rivet hole or bearing on the rivet itself. We are uncertain since the rivet material is weaker than the structural steel.
5. We are considering reinforcing the riveted shear connections in two ways:
1. Replace old rivets with new high strength bolts.
2. Add weld around the connecting angles.

If both ways are acceptable to the contractor, which solution is more cost effective?

Question sent to AISC's Steel Solutions Center

We have gaps at our column splices and are aware that a gap of 1/16" or less can be ignored. Is there something in writing that discusses the philosophy behind this statement?

Question sent to AISC's Steel Solutions Center

Do concrete cracks over composite beams affect the composite action of the beam? How can such cracking be prevented?

Question sent to AISC's Steel Solutions Center

For the 1997 AISC Seismic Provisions, can an OMF connection be designed with panel-zone web shear as the primary source of inelastic deformation or as a way to utilize the exemption for the maximum force that can be transferred by the system?

Question on SEAINT list server

If two moment girders frame into a column joint at offset elevations (3" or so), should skewed continuity plates in the column web adjoin the adjacent girder flanges, top and bottom and each side of the web (4 total), or is it better to add flat/level continuity plates at each girder flange, (8 total)?

Question sent to AISC's Steel Solutions Center

In preparing specifications for structural steel, is it common practice to have connections designed for 50% of member capacity? If so, is this a service load or factored load? What are your recommendations on how to properly specify connection design forces?

Question sent to AISC's Steel Solutions Center

In the April 2004 issue of Modern Steel Construction, the following Q/A was published:

True or False: A shape bent about its weak axis must be checked for lateral-torsional buckling.

The answer given was:

False: Lateral-torsional buckling is a phenomenon that occurs only when rotation would produce a lower energy position for that shape. A shape bent about its weak axis is already in its lowest energy position.

Question sent to AISC's Steel Solutions Center

We're designing an OMF using the IBC 2000 in Seismic Design Category "A". FEMA 350 states that lower flange backing bars should be removed to allow identification and correction of weld root flaws. On this project, the welding of the OMF joints is being observed full-time by inspectors. The question is now being asked, why are the backing bars specified to be removed? If the root pass is being observed by inspectors as it is installed, and is certified by those inspectors as being a good weld, why do we need to remove the backing bars to look at the weld again?

As the Engineer of Record, I find these arguments for leaving the backing bars in place to be compelling, but I'm not a weld expert. Can you offer me any guidance on this issue? Could the backing bars be left in place without causing problems?

Question sent to AISC's Steel Solutions Center

Please provide some opinions and guidance as to filling weld access holes in beam and column splices with weld metal for appearance sake. The AWS D1.1 welding code does not appear to address the issue. In your experience, do you usually leave them open or fill them?

Question sent to AISC's Steel Solutions Center

We were told by a steel erector that they typically do not remove shims from column base plates after grouting. They also informed us that they do not back off leveling nuts below the base plates during the grouting process. Are these practices acceptable?

Question sent to AISC's Steel Solutions Center

In the 3rd edition LRFD Manual, page 16.1-205 (or Commentary Section E4 in the 1999 LRFD Specification) states that for built-up compression members "the connectors must be designed to resist the shear forces which develop in the buckled shape." Can you provide some guidance on how to compute that required shear force?

Question sent to AISC's Steel Solutions Center

We have developed a program to check single-angle members subjected to combined axial load and bi-axial bending per the LRFD 3rd Edition spec for Single-Angle Members. We would like to check the program against some example hand calculations.

Does AISC have any example calculations that we could use, or could you direct us to another source that may have such calculations?

Question sent to AISC's Steel Solutions Center

I was wondering what the laterally unbraced length value Lb is for a cantilever? My intuition tells me that I should use twice the actual length of the cantilever for Lb, but I don't see any provisions for it in Chapter F or Appendix F of the Specification. Does limiting the Cb value to 1.0 for cantilevers provide all that is needed, and then I would just use the actual length of the cantilever for Lb?

Question sent to AISC's Steel Solutions Center

8.3.6. Why is a fillet-weld size generally limited to 1/16" less than the material thickness when placed along the edge of a connected part?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

5.1.1. Maximum hole sizes for bolts are specified in the 1999 LRFD Specification Table J3.3. What if an actual hole dimension is between two of the values?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

6.7.1. When is paint permitted on the faying surfaces of bolted connections?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

6.7.2. Both the AISC and RCSC specifications require that paint on the faying surfaces of slip-critical connections be qualified (providing a minimum slip coefficient) or that such surfaces remain unpainted. Does this requirement apply to the surfaces under the bolt head and nut?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

2.2.7. When surface roughness for thermally cut edges does not meet the limitations in 2.2.6, how is the surface repaired?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

2.2.6. What are the appropriate roughness limitations for thermally cut edges?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

10.1.2. When a paint system is required, how should it be selected?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

3.5.1. How are tolerances determined if they are not addressed in the applicable standards?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

8.3.7. Is the weld-all-around symbol acceptable when a fillet weld must be continued out-of-plane?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

10.1.1. When must structural steel be painted?

This month's Steel Interchange is compiled from a few of the many frequently asked questions (FAQs) posted at www.aisc.org/faq. Numerical references such as "2.2.6" are references to FAQs in this feature of the AISC web site.

Is wind generally considered to be a fatigue loading?

Submitted to AISC's Steel Solutions Center

The LRFD Manual of Steel Construction (3rd edition) assumes a 3" center-to-center hole spacing for most, if not all, of the examples utilizing bolted connections. Is this 3" spacing a code requirement?

Submitted to AISC's Steel Solutions Center

We have quite a few fairly short (~3") double angles to detail and are being questioned regarding using two intermediate connectors. The 3rd edition LRFD Manual appears to require a minimum of two intermediate connectors for all lengths (the tables go down to 2"). For such a short length, it seems excessive to have two intermediate connectors. Is there a way out?

Submitted to AISC's Steel Solutions Center

How common is "stub-girder" construction?

Submitted to AISC's Steel Solutions Center

On a job where the anchor rods came up short, the contractor suggested using a threaded coupler to extend the rods. What detail can be used with a threaded coupler?

Submitted to AISC's Steel Solutions Center

On a cambered beam that has welded double angles on both ends, how should the angles be oriented in the fabrication shop? Should the angles be plumb to vertical or should they be perpendicular to the curve of camber? If they are plumb then what happens when the beam comes under load? If they are fit with the beam how do you bolt up in the field?

Submitted to AISC's Steel Solutions Center.

I am designing a six-story braced-frame building with an eccentric brace. I have very large loads. I am using 2"-diameter anchor rods of ASTM A449 material, but I have also used ASTM F1554 anchor rods on a previous project. What is the recommended material anchor-rod specification? What design criteria are recommended for designing the base for these forces?

What testing should be performed on high-strength bolts upon delivery from the manufacturer?

The HSS Connections Manual indicates that a single-plate shear connection can be used if the following HSS wall slenderness ratios are satisfied:

1. Rectangular tubes: b/t < 253/Fy0.5
2. Round tubes: D/t < 3300/Fy

Are these limitations alone enough to satisfy the requirements or must one also evaluate the wall for punching shear resistance? Do the HSS requirements also apply to ASTM A53 steel pipe? How about to large-diameter round and large-sized rectangular HSS?

The question is concerning an OCBF system as specified in the AISC Seismic Provisions (May 2002). We are designing a building with horizontal braced frames (OCBF) as well as vertical braced frames, and there isn't a continuous diaphragm on which to rely for lateral shear transfer to the vertical elements. The diaphragm is made up of horizontal braced frames, and frequently there is just a single member making up the diagonal of the horizontal braced frame that serves as the load path.

In designing a horizontal braced frame, is it a requirement to still design the connection of the brace for the full tensile capacity of the brace (RyFyAg) or is it possible to consider the member a collector and just design for Ω times the force in the brace?

Is it required or recommended that the inside corners of blocked flanges (where beam flanges must be trimmed back to fit between column flanges) have radii similar to vertical web coping?

When designing a weak-axis moment connection (a beam connecting into the web of a column), where can I find information or an example calculation for this type of connection?

What is the force required to restrain a compression member from buckling laterally?

Can a partial-joint-penetration groove weld be used when it is necessary to develop the strength of a welded element?

Section 3.3 of the 2002 AISC Code of Standard Practice reads, in part, "When discrepancies exist between scale dimensions in the Design Drawings and the figures written in them, the figures shall govern." This has been interpreted to mean that if no figures exist then it is proper to scale the drawing and use the resultant dimension to build. Is this, in fact, the intent of the code?

Why did AISC stop publishing the usual gage dimension for W-shapes in the properties table? What was the basis for these values when they appeared in earlier editions (pre 9th edition ASD)?

I know there are a limited number of W-shapes considered non-compact (per ASD design). Do you have a listing of the non-compact sections for Fy = 36ksi and Fy = 50ksi?

Why is visual inspection so important in welding?

Page 4-150 of the 9th Edition of AISC ASD Manual has a table of sleeve nuts and it states "Strengths are greater than the corresponding rod when same material is used." What is the ASTM designation of these sleeve nuts?

Minoru "Ike" Ikeda, S.E.
Martin & HBL

What is the primary difference between ASTM A325 and A490 bolts, other than the difference in strength?

Can weld cracking be prevented simply by specifying notch-tough weld metal?

What constitutes "reuse" of ASTM A325 and A490 high-strength bolts? It has been my understanding that "reuse" would be a bolt that has been brought to specified pretension a second time for size and grade.

Question sent to AISC's Steel Solutions Center

In Section J3.10 of the 1999 LRFD Specification, how does one determine if deformation is a design consideration? Does this "deformation" refer to slip, i.e., does equation J3-2a apply only to slip critical connections? We are evaluating an existing building and there are no notes or details to indicate that we are dealing with slip-critical connections. Can I assume the use of equation J3-2b (which gives a higher value)?

Question sent to AISC's Steel Solutions Center

One of our observations is that the criteria φF_uA_e > R_yF_yA_g(Design Strength  > Required Strength) often cannot be satisfied with an ASTM A500 HSS anywhere along the whole length of the brace, even where there are no holes or slots. It seems to me the intent is to meet the criteria at the connection (not to build-up the section along its entire length.) Are HSS suitable for use as braces in concentrically braced frames?

Question sent to AISC's Steel Solutions Center

My question is relative to seal welding on a typical painted industrial project with the steel exposed to the environment. If there are no requirements for seal welding of all unwelded joints on a project in either the contract documents or the bid drawings, what should be the presumption made by the fabricator?

Question sent to AISC's Steel Solutions Center

The 2000 RCSC bolt specification in the 3rd Edition LRFD Manual defines sufficient thread engagement as having the end of the bolt extending beyond or at least flush with the outer face of the nut. That is easy to determine for ASTM A325 and A490 bolts. However, how would this apply to ASTM F1852 twist-off-type tension-control bolts?

Question sent to AISC's Steel Solutions Center

How can rod-type concrete anchors be welded to embedded plates?

Question sent to AISC's Steel Solutions Center

What are the damaging effects of explosions to structures?

Reprint of FAQ 12.1.4 at www.aisc.org/faq

What are some design considerations for using ASTM A588 weathering steel? Should the steel be painted?

Question sent to the AISC Steel Solutions Center

With respect to Equation A-K3.1 in the 1999 AISC LRFD Specification, a question has come up as to why the design-stress range Fsr value has to equal or be greater than the life-stress range Fth? Would it seem that Fsr. should be less than or equal to Fth? Can you clarify?

Question sent to AISC's Steel Solutions Center

How many passes does it take to make a 3/8" fillet weld? I want to tell the Engineers in our office to limit fillet-weld sizes to 5/16" (except where larger welds are absolutely needed) and I would like to show them that while a 3/8" fillet weld has 20% more strength than a 5/16" fillet weld, it takes ___% more time to install. I'm assuming that a two-pass weld takes twice as long as a one-pass weld, etc.

Question sent to AISC's Steel Solutions Center

Where can I find published information regarding the requirements for completeness of structural design drawings?

Question sent to AISC's Steel Solutions Center

During my review of the 3rd Edition LRFD Manual, I noticed that there is no mention of the T-stub moment connection as a fully restrained moment connection. Such a connection uses tee sections to connect the top and bottom flange of the beam to the column through bolting (single shear on the beam and tension with prying on the column flange). I also am aware that the extended end-plate appears to have taken the place of this bolted connection. Is AISC discouraging the use of the T-stub connection for FR moment connections?

Question sent to AISC's Steel Solutions Center

11/03/2003

There are several changes related to R (response modification coefficient) values, for example, the R value for Ordinary Steel Moment Frames changed from 4.0 (IBC 2000) to 3.5 (IBC 2003). What are the major reasons for these changes of R values?

Uksun Kim, Ph.D.
Georgia Institute of Technology

Reprint of FAQ 11.3.3 at www.aisc.org/faq

At what temperature does a typical fire burn?

Please indicate industry standards for each of the following:

1. Who is responsible for the design of the moment connections once the design criteria is provided on design drawings (reactions, moments, etc.)?
2. Should we request in our Specifications that structural calculations be provided as part of the shop drawings from the fabricator?
3. If the fabricator does not provide standard AISC details for moment connections, who is responsible for showing that they meet the design criteria?

Question sent to Editor at Structural Engineer magazine

Reprint of FAQ 11.1.7 at www.aisc.org/faq

Can spray-applied fire protection be applied to painted or galvanized steel?

Reprint of FAQ 11.2.4 at www.aisc.org/faq

Can steel continue to be used in a building after it has been in a fire? How can you assess the capacity of steel that has been exposed to fire? Are there concerns about internal or residual stress effects that have to be considered?

I am designing a steel-framed building per the 2000 IBC using ASD for steel design, so I am looking at Part III of the 1997 AISC Seismic Provisions. My question is specifically regarding the factor to apply to E in equations 4-a and 4-b.

The AISC Seismic Provisions commentary discusses the factor and from that it seems to me that I should use E/1.4 but the commentary specifically references only equations 4-1 and 4-2 regarding the factor to apply to E. E is not directly in equations 4-1 and 4-2 and it seems that the logic of using E/1.4 would apply to all ASD equations.

As I see it, the E/1.4 is simply because the IBC seismic forces are based on limit states and I am designing to ASD, so logically I should be reducing the "strength level" seismic forces to "service level" seismic forces by dividing by 1.4. Am I looking at this correctly?

Question posted on the SEAINT.ORG list server

In the 3rd Edition LRFD Manual, Example 10.11 computes the required supported beam web thickness based on bearing strength without eccentricity. Since the current design criteria developed by Astaneh for single-plate shear connections assigns eccentricity to the bolts for shear, this seems somewhat unconservative. Why is eccentricity not considered for the bolts in the bearing check?

Question sent to AISC's Steel Solutions Center

In Example 12.4 of the 3rd Edition LRFD Manual pertaining to the design of a four-bolt unstiffened extended end-plate FR moment connection, ASTM A325 slip-critical bolts were specified. Shear is transferred through slip resistance offered by surfaces of end plate and column flange. Is this the reason why full tension capacity of the bolt is used per section J3.6, instead of capacity under combined shear and tension condition per J3.7?

Is it mandatory to use slip-critical bolts instead of pretensioned bolts for this type of connection? What about a snug-tightened arrangement for this type of connection?

Question sent to AISC's Steel Solutions Center

Please advise as to where you can find the value for I_w for single-angles for equal and unequal legs. Alternatively, if not directly given by AISC, how can one calculate this property? Is it a trigonometric function of I_z?

Question sent to AISC's Steel Solutions Center

Table 4-8 in the AISC HSS Connections Manualcontains support conditions characterized as rigid and flexible for single-plate shear connections. Please explain what flexible and rigid supports imply?

Question sent to AISC's Steel Solutions Center

In the August 2002 issue of Modern Steel Construction issue you describe in Steel Interchange how anchor rods are threaded. Would the same hold true for tension-rod bracing?

Question sent to AISC's Steel Solutions Center

What are the maximum and minimum curved radii of HSS and W-shapes?

Question Sent to AISC Steel Solutions Center

I have a project where there is a need to pretension the column anchor rods to 110 kips. I am using 1.75" diameter ASTM F 1554 Grade 105 rods. The contractor proposes using a hydraulic torque wrench in lieu of a hydraulic bolt-pretensioning device due to technical, safety, schedule and cost considerations. If we were to allow him to use the hydraulic torquing, what is the necessary torque (ft-lb) required to obtain the 110 kip pretension force? In your opinion at this 110 kip tension, how much of a risk is there for the anchor rod to fail as a result of the combined tension and torsion stresses? If this was a standard slip critical bolted connection and the bolt failed during tensioning it would not be a big deal, a new bolt could easily be inserted. However with a concrete-embedded anchor rod it would be an issue if the rod were to fail.

Question sent to AISC's Steel Solutions Center

I have a specification section that refers to AISC S335. Can any one tell me if this is a valid number? I have checked the AISC web site and it does not seem to be there.

Question posted at steel-detail@yahoogroups.com

What procedures are available to repair a cracked weld between a diaphragm connection plate and the web of a box girder? The crack is more than 1" long and the web plate length is 60".

Question Sent to AISC Steel Solutions Center

Are fillet welds stronger when loaded transversely than when loaded longitudinally?

Are there any special requirements if we decide to use hot-dipped galvanized ASTM A325 Type 1 (medium carbon) bolts?

Question sent to AISC's Steel Solutions Center

When we performed the analysis of a frame, we found that our computed stresses were considerably higher than those computed by the design engineer. Upon further investigation, we found that these higher moments were due to second-order effects. The design engineer claims that he is not required to perform a second-order analysis under the ASD Specification. Is this correct?

Question sent to AISC's Steel Solutions Center

The 1999 LRFD Specification, Section I3.5b states, "Where there is only a single stud placed in a rib oriented perpendicular to the steel beam, the reduction factor of equation I3-1 shall not exceed 0.75." Does this requirement apply to ASD design as well?

Question sent to AISC's Steel Solutions Center

Can fabricators shop splice scrap shapes to make them longer? Are these splices regarded as connections to be approved by the EOR on the shop drawings?

Question sent to AISC's Steel Solutions Center

What surface preparation should be specified for steel that is to remain unpainted?

Question sent to AISC's Steel Solutions Center

Are there any differences between steel grades ASTM A572 Gr. 50 and ASTM A992?

Question sent to AISC Steel Solutions Center

What are the maximum and minimum curved radii of HSS and W-shapes?

Question sent to AISC Steel Solutions Center

Please refer to 3rd Edition LRFD Manual, Example 9.1 on pages 9-17 thru 9-20. How can the critical buckling stress fFbc be greater than 0.9Fy? In the example (Fy = 50 ksi), fFbc = 77.3 ksi, which is much greater than 0.9 x 50 = 45 ksi. I am accustomed to using f for flexure of 0.9 (when in combination with Fy), so why is this different? On page 9-7, under "Local Buckling," the wording is: connection elements are thick enough that local buckling will not limit the design strength for flexure. Does this mean that a "cap" of 0.9Fy does not apply?

Question sent to AISC Steel Solutions Center

Beams coped at both flanges are constrained to c < 2,d and d_c <0.2d in the 2nd Edition LRFD Manual, Vol. II, where c is the length of cope, d is the depth of the beam, and d_c is the depth of cope. How can beams with deeper copes ( d_c < 0.2d) be designed?

Question sent to AISC Steel Solutions Center

I often detail oversized (OVS) holes at expansion anchors as requested by the erector to facilitate field drilling. I also sometimes use OVS holes at lightly loaded connections that might be difficult to align in the field. The RCSC Specification for Structural Joints Using ASTM A325 or A490 Bolts calls for hardened washers at OVS holes, plus, the connection has to be designed as slip-critical (which requires no paint at the connection area).

1. Should I specify hardened washers on the plan and detail sheets, or is this something that the erector knows to do?
2. Do I need to detail no paint areas around the connection outer plies?

Question from the steel-detail@yahoogroups.com list-server

Referring to LRFD Specification Sections F2.2, Appendix F2.2, and Appendix G.3:

For all the standard rolled W- shapes, is the h/tw ratio always £ 260? In other words, if a standard rolled shaped is being considered, is it necessary to check for the limit states of web shear yielding or bucking? Also, for all the standard rolled W- shapes utilizing up to 50 ksi specified minimum yield strength, is it always true that:
h/tw is less than or equal to 418/SQRT(Fy)?

Stephen Crockett
D. M. Berg Consultants, P.C.

We have a customer that is interested in pursuing LEED certification on their building. They also like the idea of using recycled materials and would like to have some type of certification to the recycled content of steel. Do you have any information I can pass on to this customer?

Question sent to AISC Steel Solutions Center

My question involves the method used to produce an anchor-rod hole. Our preferred method is to simply flame-cut the 2 5/8" hole. However, the engineer is requesting that we flame-cut an undersized hole and then ream to the final diameter. This takes additional time and effort. Can you provide any documentation that would help establish a case for flame-cutting without reaming?

Question sent to AISC's Steel Solutions Center

We are unsure if we need to use a design wall thickness of 0.93t for electric-resistance-welded HSS if we use the 1989 ASD Specification. We do know that it is a requirement in the 2000 LRFD HSS Specification, but is it required in ASD?

Question sent to AISC's Steel Solutions Center

Do the limitations on maximum size of fillet welds in Section J2.2b of the 1999 LRFD and 1989 ASD Specifications apply to details such as welds on column base plates, shear single-plate shear connections, etc.

Question sent to AISC's Steel Solutions Center

09/01/2002

Is it appropriate, as stated by Ackroyd in his 1990 Engineering Journal paper, to use a representative connection stiffness equal to 50% of the initial connection stiffness for distribution of the flexible end moments for design? Are there any recommendations for an appropriate percentage of initial connection stiffness to be used when performing drift calculations for PR frames?

Question sent to AISC's Steel Solution Center

What is the slip factor between two hot-dip galvanized steel faying surfaces?

Contributed by the American Galvanizers Association (www.galvanizeit.org)

Does the AISC Specification actually allow the plate girder web to buckle (experience a slight lateral displacement) in order to develop true tension-field action?

Question sent to AISC's Steel Solutions Center

What torque do I need to achieve pretensioned ASTM A325 and A490 high-strength bolts?

Question sent to AISC's Steel Solutions Center

Is there an AISC Specification for torsion analysis per ASD?

Question posted on the SEAINT.ORG list server

What is "cold" galvanizing?

Does the galvanized steel coating of zinc resist abrasion?

Why do galvanized steel appearances differ from project to project and galvanizer to galvanizer, and is there any difference in the corrosion protection offered by the different appearing coatings?

Is there any environmental impact when the zinc coating sacrificially corrodes? Is zinc a safe metal?

What is a G90 or A60 coating?

Can galvanized steel in service withstand high temperatures for long periods of time?

Should I be concerned when galvanized steel comes in contact with other metals?

What if the article to be galvanized is larger than the dimensions of the galvanizer's kettle? Can it still be galvanized?

What is the difference between hot-dip galvanized fasteners and zinc-plated fasteners?

What is the difference between hot-dip galvanizing after fabrication and continuous hot-galvanized sheet?

How does galvanizing protect steel from corrosion?

Why would you want to paint over galvanized steel?

How well does galvanized steel perform in permanent water immersion?

Is a salt spray test in a laboratory appropriate to estimate the corrosion rate of zinc coated steel?

What are the specifications governing hot-dip galvanized steel?

Can I specify how much zinc to put on the steel?

What are the steps in the galvanizing process?

What causes wet-storage stain and how can it be prevented?

Maximum hole sizes for bolts are specified in the 1999 LRFD Specification Table J3.3. What if an actual hole dimension is between two of the values?

Question sent to AISC's Steel Solutions Center

Is there any documentation regarding the assumption an effective length factor of K = 1.0 if a P-delta analysis is performed?

Question sent to www.seaint.org

On page 6-17 of the HSS Connections Manual, a "rule of thumb" for Lw is that it should be at least equal to the HSS depth. Is this a requirement? The 1999 LRFD Specification Section J2.2b (last paragraph on page 16.1-54 of the 3rd edition LRFD Manual) requires a similar minimum distance for longitudinal welds at the end of a flat bar member in tension. However, I don't consider a tube to be a flat bar, but obviously from the wording in the HSS Manual, it seems that I should be still adhering to this rule of thumb. Using the example I described above, is it allowed to use a length of weld less than the rule of thumb as long as the connection checks?

Question sent to AISC's Steel Solutions Center

The prying action models in the 3rd Edition LRFD Design Manual do not appear consistent with those found in the 9th Edition ASD Design Manual. In particular, the b dimension has traditionally been defined as the distance from the face of the tee stem, or face of the outstanding angle leg, to the centerline of the bolt hole. However, the latest Design Manual defines the b dimension as terminating at the centerline of the outstanding angle leg. Why the change for the prying action of double-angles?

Question sent to AISC's Steel Solutions Center

Regarding the seismic force reduction factor R, there seems to be some discrepancies in values to use between the 1997 UBC and Supplement No. 2 of the AISC Seismic Provisions for Structural Steel Buildings. For example, for ordinary moment frames (OMF), the 1997 UBC requires R = 4.5 while Supplement No. 2 indicates R = 3.5. Which is correct?

Question sent to list server at www.seaint.org

FEMA 350 recommends a special weld access hole configuration for use with certain moment connections for ordinary and special moment frames. It is my understanding that this weld access hole configuration will also be included in the 2002 AISC Seismic Provisions. As you know, there has been an allegation that use of this weld access hole configuration constitutes an infringement of the patent for a proprietary slotted-web moment connection. What is AISC's position on this issue?

C. Mark Saunders, S.E.
Rutherford and Chekene Engineers
San Francisco, CA

An inspector on our job insists that A36 is no longer manufactured. I have recently taken an advanced steel course and heard no mention of such a fact. I understand that the A572 Specification has been refined, but I find it difficult to believe that A36 is gone altogether. If you could direct me to a reference I would greatly appreciate it.

Question sent to AISC Steel Solutions Center

Why has the height limitation of 160 ft for OMFs in UBC 97 been reduced to 35 feet in the IBC 2000, for structural steelwork buildings in Seismic Design Category (SDC) D? I can't point to an exact reason, but commentary from some of the steel seismic seminars leads me to believe that AISC wants people use special frames of all types for almost everything (except maybe SDC A and B). I would expect the penalties to keep going for using ordinary frames in zone with moderate seismicity as well.

Are there any types of moment resistant HSS (beam) to HSS (column) connections that could satisfy the requirements for Special Moment Frames? Are there any types of HSS (beam) or HSS (column) to HSS (brace) connections that could satisfy the requirements of Concentrically Braced Frames?

Joaquin Fidalgo
Bogota, Colombia
South America

I have a low building in Seismic Design Category D and have run into a problem designing the bracing connection in an Ordinary Concentrically Braced Frame (OCBF). There used to be an exemption for low buildings (Section 14.5 of the 1997 Seismic Provisions) in OCBFs. It is unclear to me whether or not these exemptions still exist in AISC Seismic Provisions Supplement No. 2. The new provisions make it look as if the engineer has to design the brace connection in this low building to develop the full tensile capacity of the brace...for serviceability reasons, my brace is a 6'6 HSS (with a tensile capacity of about 300 kips). However, the greatest design load on the brace is only about 30 kips. Does the bracing connection have to be designed for the tensile capacity of the brace? Is there a good way out of this?

Question sent to AISC Steel Solutions Center

The AISC Seismic Provisions Supplement No. 2 Commentary to Sections 10 and 11 regarding IMF and OMF, respectively, states "As a result of the SAC program (FEMA 2000a), the IMF as defined in the 1997 AISC Seismic Provisions for Structural Steel Buildings is no longer applicable. This system has been eliminated and the OMF as given in AISC (1997) has been split into two systems: the IMF based on a tested design procedure and the OMF based on a prescriptive design procedure." My question is that IBC 2000 currently prescribes an R value of 4 for an OMF and 6 for an IMF. This commentary seems to suggest that the level of ductility provided by an OMF and IMF are comparable and, therefore, the R values should be approximately the same. What are the thoughts of the AISC Seismic Provisions Committee? I have read the summary of the Lehigh test data for the prescriptive connection and, if I recall correctly, the level of ductility available within this connection if the current requirements are followed (weld metal meeting CVN toughness requirements, backer bar removal and welding requirements, welded beam web, good access hole geometry, and continuity stiffener requirements) is pretty high.

Question sent to AISC Steel Solutions Center

Given the correct loading conditions, it is possible for nuts on anchor rods to pull through anchor rod holes or, when leveling nuts are used and the column is not grouted, for the base plate to push through the leveling nuts. Both failure modes may occur when a washer of insufficient size (diameter and thickness) is used. What procedures are available for properly sizing the washer to help prevent such failures?

When designing moment frames, we will typically use continuity plates between the column flanges when a beam or girder frames into the column. Although this keeps the column size down, it also increases fabrication cost. Do you have any comments regarding the elimination of continuity plates by using a heavier column?

Question sent to AISC's Steel Solutions Center

In the 2000 AISC Code of Standard Practice, Section 7.13.1.1 states "the angular variation of the working line from a plumb line shall be equal to or less than 1/500 of the distance..." Is this the angle between the working line and plumb line, or is it the ratio of the displacement to the column length?

Question sent to AISC's Steel Solutions Center

Where can I find literature or references regarding the design of composite floors with penetrations?

According to Section B2 of the 1999 LRFD and 1989 ASD Specifications, the width of a bolt hole must be taken as 1/16" greater than the nominal dimension of the hole when calculating the net area. Please clarify whether this provision should read as 1/16" greater than the nominal bolt diameter, as the nominal hole dimension tables in the manuals are based on adding a 1/16" to the nominal diameter of the bolt.

Question sent to AISC's Steel Solutions Center

ASTM A992 wide-flange shapes and F1554 anchor rods are not listed in the 9th edition ASD Manual nor in the 1989 ASD Specification. Can I use these newer materials in designs involving the ASD, or does the design need to be based on the LRFD Specification?

Comment sent to AISC's Steel Solutions Center

Section 8.3a(2) of 1997 AISC Seismic Provisions [Section 8.4a(2) in the 2002 version] states that the minimum required strength for each flange shall be 0.5RyFyAf, where Af is the flange area of the smaller column connected. Does Af pertain to the area of just one flange or both flanges of the smaller column?

Question sent to AISC's Steel Solutions Center

On page 7-114 of the AISC ASD/LRFD Connections Manual, a sketch of the Whitmore Section is shown for a particular example. Consider a connection with the brace connecting to a beam only (no column). I see the dimensions Lw, l1 and l2 remaining the same. The dimension l3 will increase in length from the corner of the section until a line parallel to the brace intersects the beam flange. Then per the last paragraph on the previous page, the column length would be the average of l3 and l1 (l2 is negative in this case). Would this be correct?

Question sent to list server at www.seaint.org

The AISC 9th Edition (ASD) illustrates procedures for bolted hanger-type connections with a single line of resistance to prying action on each side of the hanging member. If each line of resistance consists of a bolt group, what design and analysis methods should be used?

Jay Shniderman, P.E.
Van Nuys, CA

The 9th Edition ASD Manual illustrates procedures for bolted hanger-type connections with a single line of resistance to prying action on each side of the hanging member. If each line of resistance consists of a bolt group, what design and analysis methods should be used?

Jay Shniderman, P.E.
Van Nuys, CA

Symmetric tension bolt pitches are assumed in the published design procedures for this connection. However, due to ease of fabrication, we would like to use a different pitch above and below the top tension flange of the beam. Are there guidelines on this, or has this connection been prequalified for only symmetric pitches?

Is it true that I can take advantage of the 2000 LRFD HSS Specification for structural designs involving ASTM A53 Grade B pipe? What, then, are the differences between HSS and and pipe if both use the same specification?

In exterior columns, should the eccentricities resulting from the beam connections be considered when the connections are not designed by the SER? For example, when a W-shape beam is framed into an exterior HSS column via a single-plate shear connection, the column could be designed for the eccentricity equal to the distance from its centerline to the bolt line. With that approach, it might make sense not to place the beams at the column lines and laterally brace the column by a light angle section.... Alternatively, the beams could be assumed to extend into the column centerlines and the specialty connection design engineer directed to design the connection for combined shear and moment. Can the AISC ASD Manual's tables for single-plate shear connections or eccentric bolted connections be used for that purpose?

If the eccentricity is considered in column design, it should presumably be applied in two directions in corner columns and in columns where the exterior girders deliver vastly unequal reactions from the opposite sides. This might lead to the corner columns actually being heavier than the interior columns, which support four times the load.

Alexander Newman, P.E.
Maguire Group Inc.
Foxborough, MA

Does AISC or another organization publish specific standards or specifications for steel detailing and shop drawings? The drawings that I've seen coming from some of the new CAD software have not been consistent from one job to the next, nor have they matched the clarity that good steel detailers produce by hand. NISD publishes guidelines for information to be shown by the design engineer but nothing on standards for what the detailers will provide. I'm looking for some good balanced standards to reference as minimum requirements for steel shop drawings that are submitted to us. Some detailers have advised me that the information is there in the software. So, how can I communicate my requirements up front, so that the advantages of electronic data transfer are realized and properly balanced with the need for clear record documents?

Richard A. Meloy, P.E.
Butler Heavy Structures
Kansas City, MO

I have a question on designing single-plate shear connections. Should the weld between the plate and support be designed for shear only, or for both shear and bending moment?

Question sent to AISC's Steel Solutions Center

How does one design a structural WT member under flexure when the stem is in compression? Chapter F of the 1989 ASD Specification does not appear to address this particular case?

In order to check the minimum gusset plate thickness against the fillet weld size required for strength, the 3rd edition LRFD Manual contains the expression t_min = 6.19D/F_u. However, in the 2nd edition LRFD Manual, the expression is t_min = 5.16D/F_y. Is the new expression based on F_y = 0.833F_u and why use F_u instead of F_y?

Question sent to AISC's Steel Solutions Center

The 1985 RCSC Bolt Specification found in the 9th edition ASD Manual uses the phrase "full thread engagement." However, the 2000 RCSC Bolt Specification appears to have dropped this nomenclature. Is the phrase "full thread engagement" still used by the RCSC Bolt Specification, and, if so, where can we locate this information?

Question sent to AISC's Steel Solutions Center

I would like to design circular column base plates. However, there appears to be little or no information on the subject. Does anyone know of papers, articles or design guidelines for the design of circular base plates?

Question sent to AISC's Steel Solution Center

I am preparing to design a four-story moment frame office building on the east coast (seismic zone 0) and was wondering what other folks have been doing. What are some other engineers in this area requiring? What is standard practice?

Question sent to www.seaint.org

Is it appropriate, as stated by Ackroyd in his 1990 Engineering Journal paper, to use a representative connection stiffness equal to 50% of the initial connection stiffness for distribution of the flexible end moments for design? Are there any recommendations for an appropriate percentage of initial connection stiffness to be used when performing drift calculations for PR frames?

Question sent to AISC's Steel Solution Center

ASTM F1554 Grade 105 anchor rods will be used in a coastal area with wind gusts of 130 mph located in a very high seismic zone. We are planning to pretension the anchor rods to avoid the risk of inducing tensile fatigue from loading cycles resulting from wind loads. Do you agree with this as being a valid reason to pretension these rods?

Question sent to AISC's Steel Solution Center

Is there an AISC spec for painting at moment connections or is our erector just complaining about his welders having to burn through paint to make field welded moment connections?

Question sent to the steel detailers list server at Yahoo Groups http://groups.yahoo.com/group/steel-detail/

A supplier on one of our projects is claiming that when we specify ASTM A307 anchor rods, he can supply anchor rods with rolled threads. The motivation is that these rods are slightly cheaper. His argument is based on the claim that ANSI B1.1, which is referenced by A307, reportedly lists rolled threads as an option. This causes problems since the bolt shank is slightly smaller which results in a weaker bolt and results in a larger space between rod and hole. I believe that the supplier is wrong since A307 specifies a minimum area and a tensile capacity that "trumps" B1.1. I may be wrong if the minimum area is measured at the threaded region. Is the supplier blowing smoke or do we need to modify our specifications to exclude rolled threads?

Question posted at SEAINT list server www.seaint.org

With regards to built-up shapes, what are the minimum and maximum web thickness and flange thickness? Is there a minimum or maximum relationship between t_f and t_w?

Question sent to AISC Steel Solutions Center

Could anybody advise me on how to check the adequacy of galvanization of members of steel towers used for power transmission lines? What are the specifications that apply, and is the thickness of galvanization proportional to the thickness of the member?

Question sent to AISC Steel Solutions Center

A rigid support forces any beam-end rotation to take place within the connection. This leads to horizontal forces in the bolts, creating a couple. It seems to me that this imposed horizontal force in the bolts and plate will reduce the available capacity in a vertical direction when vectorially subtracted from the bolt's capacity (or plate's bolt-bearing strength, or weld capacity). Yet the HSS Connections Manual Table 4-8, and the LRFD Manual Table 9-10 (identical tables) show rigid connections having more shear capacity than flexible ones. Why is this?

Question posted at SEAINT list server (www.seaint.org)

Please explain how workable gages are calculated in the 3rd edition LRFD Manual. The dimension tables in the new manual sometimes use the same workable gage for flange widths that differ considerably for a given beam depth, such as W14. Why is this?

Question sent to AISC Steel Solutions Center

I am designing using the ASD Specification. Over the past few years, I have seen ASTM A992 specified for wide flange shapes and ASTM F1554 for anchor rods. These materials are not listed in the ninth edition ASD Manual. Can I design with these materials if I use ASD?

Question sent to AISC's Steel Solutions Center

Can I use the fillet encroachment table in the AISC Manuals for encroachments to the web instead of those to the flange?

Question sent to AISC's Steel Solutions Center

The Society for Protective Coatings (SSPC) specifies four different levels of blast cleaning: SP5, SP6, SP7, and SP10. Which form of blast cleaning must be used to obtain a Class B uncoated surface for a slip critical connection? Similarly, is there a minimum angular surface profile required for a Class B surface? Also, if blast cleaned steel develops surface rust while being stored on site, will it still meet the Class B requirement?

Question sent to AISC's Steel Solutions Center

Does anyone know what to do when you have a moment connection with a "thick" plate on top of the upper beam flange welded to it and to the column and you have to place the steel deck? Do I place the steel deck on top of it (it doesn't look good from the lower floor or level) or do I cut the steel deck around the perimeter of the plate so that the deck would rest on the upper flange level of the beam?

Question sent to www.seaint.org

When X-bracing is used in a situation where the two braces share the load, can the brace in tension be counted on to laterally brace the compression strut at midpoint against out-of-plane buckling (assuming the two braces are connected at midpoint)?

Question sent to AISC's Steel Solutions Center

One of the issues in determining fire resistive properties per ASTM E 119 is whether the members are considered restrained. It is my understanding that the steel industry has decided that all moment connections result in a restrained condition. Does anyone know if there is any documentation on this?

Question posted on SEAINT list server

Question #6 in the March 2002 Steel Quiz stated that the reduction coefficient U for shear lag is applied to the net area of bolted tension members and to the gross area of welded tension members except for HSS members with slots for gusset plates. What is meant by this exception?

Question submitted anonymously

What reference material is available for the design of square tubular section arches with respect to in-plane and out-of-plane buckling? Loading may be full uniform, partial uniform, or concentrated load at quarter or half point. Arches are braced laterally at third points for typical spans. Most technical papers that I have reviewed are concerned with I shape sections although some of the analysis is transferable.

Susan Guravich, P. Eng.
Skarborn Engineering Ltd.

Often we hear that bolts should be installed pointing downward (with the heads on top) to prevent them from falling off if the nut got loose and slipped. Is there any code or steel installer manual or any written reference that supports, recommends or mandates this practice?

Question sent to AISC's Steel Solutions Center

Is it accurate to state that specifying a connection as slip-critical will increase the strength of that connection?

Karl Lankenau
Dormitory Authority of the State of New York
Albany, NY

I heard that AISC has information for designing composite beams with a cover plate on the bottom flange. Please let me know how I could obtain this information.

Question sent to AISC's Steel Solutions Center

My question is regarding compression buckling of gusset plate attached to a beam and column. For example, if my brace is in compression, how do I check the KL/r for the effective section of the gusset plate?

Question sent to AISC's Steel Solutions Center

Please provide any additional sources of information relative to AISC Specification section E4 (ASD Manual, 9th ed., pages 5-43 and 5-44) covering the proportioning of lacing members to resist a 2% shear stress.

Additional references with discussions, the history of its origin, derivation and/or examples of the proper application of this specification would be greatly appreciated.

W. H. Parker
Derrick Engineering

How is a masonry shelf-angle designed?

Question reprinted from www.aisc.org/faq.html

When should backing bars and run-off tabs be removed after welding?

Frequently asked question on AISC website, www.aisc.org/faq.html

What is the difference in design philosophy between a building structure that has been designed to meet the AISC LRFD Specification for Structural Steel Buildings and a building that has been designed to meet the AISC Seismic Provisions for Structural Steel Buildings?

Frequently asked question on AISC website, www.aisc.org/faq.html

Shall I use the load combinations (shown below) with amplified earthquake load (Eqs. 4-1 and 4-2 of the 1997 AISC Seismic Provisions for Structural Steel Buildings) for the design of column base plate and anchor rods?

1.2D + 0.5L + 0.2S + Omega₀Q_E
0.9D - Omega₀Q_E

Mike Ginsburg, P.E.
Leo A. Daly
Omaha, NE

I have a situation where I would like to use some relatively small diameter high strength (A325 or A490) bolts. What is the smallest diameter bolt that I can specify?

Submitted anonymously

Who manufactures equipment for cambering steel?

Cheryl Vickroy
Research Fellow
Madison, WI

Are there any recommended procedures for designing single plate shear connections requiring multiple lines of bolts or a deep single line of bolts?

Question sent to Steel Solutions Center

What is the difference between traceability and identification of material?

Reprinted from www.aisc.org/faq.html

In the 1992 and 1997 Seismic Provisions, for SMF, the resistance factor for panel-zone web shear is 0.75. The Seismic Provisions are somewhat silent for panel-zone web shear in OMF. LRFD Specification Section K1.7 using a resistance factor for panel-zone web shear of 0.90. For OMF, do we default to Section K1.7 and use a resistance factor for panel-zone web shear of 0.90? Or is the resistance factor always 0.75 in OMF and SMF if the loading is non-static?

Stephen Crockett
D. M. Berg Consultants, P.C.

What are some good references for designing curved structural members?

Question sent to AISC's Steel Solutions Center

One rule of thumb that our office uses for the minimum projection of bolts beyond the nut is 4.5 threads. Is there any written information documenting a minimum allowable projection?

Question sent to AISC's Steel Solutions Center

In the 1989 ASD Specification, Section F1.3 provides calculations for allowable stresses in members with unbraced lengths greater thanL_c. Equation F1-8 seems to provide a very low allowable stress for these members, yet it can be applied to members for any value of l/r_t. Is there something that I am overlooking?

Question sent to AISC's Steel Solutions Center

Is there any situation where the design of a slip critical bolted connection would not be required to address bearing requirements? How about a connection using slotted holes?

Question sent to AISC's Steel Solutions Center

When calculating biaxial bending in beams, we generally use ASD equation H1-3:f_a/F_a + f_bx/F_bx + f_by/F_by < = 1.0 which in many cases, allows us to use 0.66F_y as F_bx and 0.75F_y as F_by. However, we've recently reviewed calculations where F_bx and F_by are 0.60F_y, no matter what the compact length attributes of the beam are. Is the 0.6F_y for biaxial bending standard?

Question from SEAINT list server www.seaint.org

In South Carolina we are using the provisions of IBC 2000 and most of the state's structures are now under seismic design category D. Section 2212.1.2 requires the use of AISC's Seismic Provisions for Structural Steel Buildings, dated April 15, 1997. If we design using Part III of this document (page 135, Part III Section C4.1), it encourages the use of the governing code or standard for the load factor on E. This governing code is now IBC 2000. If we use the alternate IBC 2000 load combinations of section 1605.3.2, is the intent that the combinations of formula 16-17 and 16-18 be used now instead of the equations 4-a and 4-b in AISC's Seismic Provisions? Please verify if the nominal strength per section 4.2a is to be used for equations 4-a and 4-b and 4-1 and 4-2 that it is ok to use this nominal strength with formulae 16-17 and 16-18 of the IBC 2000 (and not use any other increase factors). Paragraph 1605.3.2 of IBC 2000 states that stresses are permitted to "be increased where permitted by .... the referenced standard." What increases do the steel standards (seismic) allow for the formulae in section 1605.3.2? The IBC 2000 says in section 1617.1.2 to use an increase of 1.7 and Phi of 1.0 in ASD for E_m in equations 16-30 and 16-31. Do these take the place of formulae 4-1 and 4-2 in AISC's Seismic Provisions, or should this be checked and also 4-1 and 4-2 with the actual resistance factors for steel as per section 4.3 on design?

Question sent to AISC's Steel Solutions Center
Greenville, SC

Can headed studs be welded thru 18 ga. metal deck without having the deck pre-punched?

Question from SEAINT list server www.seaint.org

There are numerous sources that provide recommendations and opinions regarding permissible lateral drift of steel buildings that are supporting exterior walls comprised of brick veneer or concrete masonry unit (CMU) block. These include AISC Design Guide No. 3 Serviceability Design Considerations for Low-Rise Buildings by J. M. Fisher and M. A. West. Does any other established entity comparable to AISC provide explicit specifications for this situation?

Kevin B. Westervelt, S.E., P.E.
Knoxville, TN

It is a general rule that welding on an existing structural member is not permitted unless provisions are made to unload the member first (for example, if the member is being reinforced) and that the weld must not degrade the properties of the material.

Is there a written reference that discusses this, both from a code perspective, and a practical approach?

Alan L. Blosser P.E.

Are there any guidelines or recommendations concerning the repair of anchor rods without adequate projections? This question applies particularly to applications in rigid frames and braced frames where tension is a limiting design condition. Also these are applications where epoxy anchors are not applicable. We know of several methods of repair - couplers or cutting and welding bolt projections. Could you supply some information on the applicability of each repair-minimum/maximum size of anchor, minimum/maximum projection, minimum/maximum plate size?

Kurt Swensson
KSI Structural Engineers
Atlanta, GA

While researching a column plumbness issue on a recent project, a dispute arose over the definition of the "building line" shown in Figure C-7.5 of the 2000 Code of Standard Practice for Steel Buildings and Bridges. We argued that the building line represented the exterior of the building, therefore at the 20th floor the column must not extend more than 1" towards the exterior or 2" towards the interior of the building. Where can we find AISC's definition of the term "building line" and are we correct in our interpretation?

Greg Ruberto, P.E.
Civilsmith Engineering, Inc.
State College, PA

I would like to know if there's a reference similar to what could be called "Crane Rail Design for Dummies." I am looking for a handbook or manual on cranes for industrial buildings. Subjects mainly needed are bridge beams, runway beams, columns, bracings, brackets, load considerations, load combinations, etc. Crane information is not relevant because the supplier or manufacturer supplies it all. I would like a reference that focuses on the structure that supports them.

Question from Structural Engineers Association International (SEAINT) email list-server

Due to architectural characteristics, I am in the process of plan-checking a few moment frames using curved members (curved beams to column). I have questions regarding curved moment frames.

1. Are they allowed in current codes?
2. Has there been testing done on curved moment frames?
3. In calculations, how do you design the beam-column connection? Will torsion be introduced in this connection and the frame members?
4. If HSS beams and columns are used, please suggest a beam to column connection that will be acceptable.

David Chung, P.E.

Regarding Chapter K of the ASD Manual, reference page 5-82 section K1-8 paragraph 3:

If Sections K1.4 or K1.6 requires stiffeners, the stiffeners shall be designed as axially compressed members (columns) in accordance with requirements of Section E2.

If Section K1.4 requires the stiffener, I would design the stiffener as compression member with an axial load of R from section K1.4. If Section K1.6 requires the stiffener, should the stiffener be designed for an axial load of P_bf from section K1.6 or from the computed force delivered by the flange? If P_bf is used, often the stiffener (assuming the same width as the flange) will be thicker than the flange and this appears odd to me.